NEHRP Recommended Seismic Provisions Cover NEHRP (National Earthquake Hazards Reduction Program) Recommended Seismic Provisions for New Buildings and Other Structures (FEMA P-750) 2009 Edition Prepared for the Federal Emergency Management Agency of the U.S. Department of Homeland Security By the Building Seismic Safety Council of the National Institute of Building Sciences BUILDING SEISMIC SAFETY COUNCIL A council of the National Institute of Building Sciences Washington, D.C. 2009 NOTICE: Any opinions, findings, conclusions, or recommendations expressed in this publication do not necessarily reflect the views of the Federal Emergency Management Agency. Additionally, neither FEMA nor any of its employees make any warranty, expressed or implied, nor assume any legal liability or responsibility for the accuracy, completeness, or usefulness of any information, product, or process included in this publication. The Building Seismic Safety Council (BSSC) was established in 1979 under the auspices of the National Institute of Building Sciences as a forumbased mechanism for dealing with the complex regulatory, technical, social, and economic issues involved in developing and promulgating building earthquake hazard mitigation regulatory provisions that are national in scope. By bringing together in the BSSC all of the needed expertise and all relevant public and private interests, it was believed that issues related to the seismic safety of the built environment could be resolved and jurisdictional problems overcome through authoritative guidance and assistance backed by a broad consensus. The BSSC is an independent, voluntary membership body representing a wide variety of building community interests. Its fundamental purpose is to enhance public safety by providing a national forum that fosters improved seismic safety provisions for use by the building community in the planning, design, construction, regulation, and utilization of buildings. This report was prepared under Contract HSFEHQ-04-C-0465 between the Federal Emergency Management Agency and the National Institute of Building Sciences. For further information on Building Seismic Safety Council activities and products, see the Council’s website (www.bssconline.org) or contact the Building Seismic Safety Council, National Institute of Building Sciences, 1090 Vermont, Avenue, N.W., Suite 700, Washington, D.C. 20005; phone 202-289-7800; fax 202-289-1092; e-mail bssc@nibs.org. Copies of this report on CD Rom may be obtained from the FEMA Publication Distribution Facility at 1-800-480-2520. Limited paper copies also will be available. The report can also be downloaded in pdf form from the BSSC website at www.bssconline.org . The National Institute of Building Sciences and its Building Seismic Safety Council caution users of this Provisions document to be alert to patent and copyright concerns especially when applying prescriptive requirements. FOREWORD One of the goals of the Federal Emergency Management Agency (FEMA) and the National Earthquake Hazards Reduction Program (NEHRP) is to encourage design and building practices that address the earthquake hazard and minimize the resulting risk of damage and injury. Publication of the 2009 edition of the NEHRP Recommended Seismic Provisions for New Buildings and Other Structures (FEMA P-750) reaffirms FEMA’s ongoing support of efforts to achieve this goal. First published in 1985, the 2009 edition of the Provisions marks the seventh in a series of updates to the document and several complementary publications. FEMA is proud to have sponsored this project conducted by the Building Seismic Safety Council (BSSC) of National Institute of Building Sciences (NIBS) and continues to encourage the widespread dissemination and voluntary use of this state-of-art consensus resource document. In contrast to the earlier editions of the Provisions which resulted from three-year update projects, the 2009 edition is the first resulting from a five-year update effort that allowed the BSSC’s Provisions Update Committee (PUC) to make some major changes in both the substance and the format of the Provisions document. The most significant change involves the adoption by reference of the national consensus design loads standard, ASCE/SEI 7-05, Minimum Design Loads for Buildings and Other Structures, including the related consensus standards referenced therein and Supplements 1 and 2. Part 1 of this document includes consensus-approved modifications of the seismic requirements in the standard. Among these modifications is the adoption of new seismic design maps based on seismic hazard maps issued in 2008 by the U.S. Geological Survey (USGS) along with some design-related adjustments. Another major change has been made to the accompanying Commentary, previously issued as a separate volume but now included as Part 2 of the 2009 Provisions. The content of the Commentary has been completely rewritten to provide users with an up-to-date, user friendly explanation of how to design using the Provisions and the reference standard. Part 3 of the 2009 Provisions consists of a series of resource papers intended to clarify aspects of the Provisions, stimulate consideration of and feedback from the design community on new seismic design concepts and procedures, and/or encourage the development and adoption of new requirements in ASCE/SEI 7 and the standards referenced therein. Thus, the 2009 Provisions serves as a national resource intended for use by both design professionals and the standards- and codes-development community in fostering development of a built environment designed and constructed to protect building occupants from loss of life and serious injury and to reduce the total losses from future earthquakes. FEMA wishes to express its deepest gratitude for the significant efforts of the over 200 volunteer experts as well as the BSSC Board of Direction, member organizations, consultants, and staff who made the 2009 NEHRP Recommended Seismic Provisions possible. Americans unfortunate enough to experience the earthquakes that will inevitably occur in the future will owe much, perhaps even their lives, to the contributions and dedication of these individuals. Without the expertise and efforts of these men and women, this document and all it represents with respect to earthquake risk mitigation would not have been possible. Federal Emergency Management Agency Page intentionally left blank. PREFACE and ACKNOWLEDGEMENTS Since its creation in 1979, the National Earthquake Hazard Reduction Program (NEHRP) has provided a framework for efforts to reduce the risk from earthquakes. The Building Seismic Safety Council (BSSC) is extremely proud to have been selected by the Federal Emergency Management Agency (FEMA) to play a role under NEHRP in improving the seismic resistance of the built environment. Further, the BSSC is pleased to mark the occasion of its thirtieth anniversary with delivery to FEMA of the consensus-approved 2009 NEHRP Recommended Seismic Provisions for New Buildings and Other Structures, the eighth edition of this landmark publication. The Provisions has evolved over the past three decades into a widely available, trusted, state-of-theart seismic design resource document with requirements that have been adapted for use in the nation’s model building codes and standards. Work on the 2009 Provisions began in September 2004 when the National Institute of Building Sciences, the BSSC’s parent organization, entered into a contract with FEMA for initiation of the 2009 Provisions update effort. Based on input from the BSSC member organization representatives and alternate representatives and the BSSC Board of Direction, individuals to serve on the 2009 Provisions Update Committee (PUC) and its Technical Subcommittees (TSs) and ad hoc Issue Teams were identified. The PUC and its TSs and ITs were fully established in 2005 as was a Seismic Design Procedures Review Group (SDPRG) charged to assess ongoing work by the U.S. Geological Survey (USGS) to update its seismic hazard maps. It is the collective efforts and expertise of the national experts serving on these groups that is reflected in the 2009 Provisions. In recognition of the fact that the codes and standards arena now operates differently than it did in the past, the format of 2009 Provisions has been changed to focus more on exploration of new technologies and procedures and less on format and editorial changes. To this end, the national consensus design loads standard, Minimum Design Loads for Building and Other Structures, ASCE/SEI 7-05 (including Supplements 1 and 2), has been adopted as the primary reference standard of the Provisions. Areas of the standard in need of modification also were identified and proposals to do so were prepared and voted on by the membership. These modifications appear in Part 1 of this document and, together with ASCE/SEI 7-05 and the references cited therein, constitute the 2009 Provisions. (A summary of the results of the member organization ballots and comment resolution process is available from the BSSC upon written request and will be posted on the BSSC website at www.bssconline.org.) A major effort also was made to rewrite the commentary to the Provisions. Until now, the commentary was published as a separate volume accompanying the Provisions and tended to explain the development of the existing requirements. For 2009, the commentary appears in Part 2 of this Provisions document and explains how to apply the Provisions requirements as articulated in ASCE/SEI 7-05 and the references cited therein. (Note that the Part 1 modifications to the standard are accompanied by their own commentary text.) Part 3 of this volume is a collection of resource papers. Included are substantive proposals on topics that require further consideration by and feedback from the seismic design community before they become Provisions requirements as well as papers that clarify some aspects of the Provisions requirements. In addition, three of the appendices from the 2003 Provisions are still considered to include information of interest and they also are included in Part 3. As in the past, the 2009 Provisions would not have been possible without the expertise, dedication, and countless hours of effort of the more than 200 dedicated volunteers who participated in the update process. The American people benefit immeasurably from their commitment to improving the seismic-resistance of the nation’s buildings. A list of all those who participated in the 2009 Provisions update project is included as the Appendix of this volume, but a few individuals and groups deserve special recognition. As Chairman of the BSSC Board of Direction, it is my pleasure to express heartfelt appreciation to: • The members of the BSSC Provisions Update Committee, especially to PUC Chair Ronald Hamburger; • The members of the Seismic Design Procedures Review Group, especially Chair Charles Kircher and Nicolas Luco of the USGS; • FEMA Project Officers Michael Mahoney and Mai Tong and FEMA Subject Matter Expert Robert Hanson; • Michael Valley who worked with the update committees to draft the Part 2 commentary to the 2009 Provisions; • The representatives of the BSSC member organizations who devoted considerable time and attention to the four individual rounds of balloting that were required to produce the 2009 Provisions document; and • The BSSC staff who work tirelessly behind the scenes to support all the update groups and who bring the finished product forward for acceptance. Finally, I wish to thank the members of the BSSC Board of Direction who recognize the importance of this effort and provided sage advice throughout the update. We are all proud of the 2009 NEHRP Recommended Seismic Provisions and it is my pleasure to introduce them. David Bonneville Chairman, BSSC Board of Direction CONTENTS Foreword ....................................................................................................................................................................iii Preface and Acknowledgements ................................................................................................................................ v 2009 NEHRP RECOMMENDED SEISMIC PROVISIONS FOR NEW BUILDINGS AND OTHER STRUCTURES: PART 1, PROVISIONS ................................................................................................................... 1 1.1 Intent ............................................................................................................................................................... .3 1.2 Reference Document ........................................................................................................................................ 3 Commentary to Sections 1.1 and 1.2 ........................................................................................................ 3 1.3 Modifications to ASCE/SEI 7-05 ..................................................................................................................... 5 Chapter 11, Seismic Design Criteria Section 11.1.2, Scope ............................................................................................................................... 5 Section 11.2, Definitions .......................................................................................................................... 5 Section 11.3, Notation .............................................................................................................................. 5 Section 11.4, Seismic Ground Motion .................................................................................................... 6 Section 11.8, Geologic Hazards and Geotechnical Investigation ............................................................. 8 Commentary to Section 11.1.2 ............................................................................................................... 10 Commentary to Section 11.2 .................................................................................................................. 10 Commentary to Section 11.4.3 and 11.4.4 ............................................................................................. 10 Commentary to Section 11.8.3 ............................................................................................................... 19 Chapter 12, Seismic Design Requirements for Building Structures Table 12.2-1, Design Coefficients and Factors for Basic Seismic-Force-Resisting Systems ................. 21 Table 12.6-1, Permitted Analytical Procedures ...................................................................................... 25 Section 12.8.7, P-delta Limit .................................................................................................................. 26 Section 12.11.2.2.1, Transfer of Anchorage Forces into Diaphragm ..................................................... 26 Section 12.11.2.2.3, Wood Diaphragms ................................................................................................. 27 Section 12.14.7.5.1, Transfer of Anchorage Forces into Diaphragm ..................................................... 27 Section 12.14.7.5.2, Wood Diaphragms ................................................................................................. 27 Section 12.14.8.1, Seismic Base Shear .................................................................................................. 27 Commentary to Section 12.6 .................................................................................................................. 28 Commentary to Section 12.8.7 ............................................................................................................... 28 Additional References for Chapter 12 Commentary .............................................................................. 28 Chapter 13, Seismic Design Requirements for Nonstructural Elements Section 13.6.5.5, Additional Requirements [for Component Supports] ................................................. 29 Section 13.6.8.2, Fire Protection Sprinkler Systems in Seismic Design Category C ............................. 29 Section 13.6.8.3, Fire Protection Sprinkler Systems in Seismic Design Categories D through F ........................................................................................................................................ 29 Commentary to Section 13.6.5.5 ............................................................................................................ 29 Commentary to Section 13.6.8.2 ............................................................................................................ 29 Chapter 14, Material Specific Seismic Design and Detailing Requirements Section 14.1, Steel ................................................................................................................................. 31 Section 14.2.2, Modifications to ACI 318 .............................................................................................. 33 Section 14.2.3, Additional Detailing Requirements for Concrete Piles and Section 14.2.3.1, Concrete Pile Requirements for Seismic Design Category C ....................... 34 Section 14.2.3.2, Concrete Pile Requirements for Seismic Design Categories D through F .................. 35 Section 14.2.4, Acceptance Criteria for Special Precast Structural Walls Based on Validation Testing .......................................................................................................................... 36 Section 14.4.5, Modifications to Chapter 1 of ACE 530/ASCE 5/TMS 402.......................................... 40 Commentary to Section 14.1.1 ............................................................................................................... 40 Commentary to Section 14.1.4 ............................................................................................................... 40 Commentary to Section 14.2 .................................................................................................................. 41 Commentary to Section 14.2.3 ............................................................................................................... 42 Commentary to Section 14.2.4 ............................................................................................................... 43 Commentary to Section 14.4.5 .............................................................................................................. 53 Additional References for Chapter 14 Commentary .............................................................................. 54 Chapter 15, Seismic Design Requirements for Nonbuilding Structures Table 15.4-2, Seismic Coefficients for Nonbuilding Structures Not Similar to Buildings ..................... 57 Section 15.5.3, Steel Storage Racks ....................................................................................................... 57 Section 15.6.2, Stacks and Chimneys ..................................................................................................... 57 Section 15.7.6, Ground-Supported Storage Tanks for Liquids ............................................................... 58 Commentary to Section 15.5.3 ............................................................................................................... 58 Commentary to Section 15.6.2 ............................................................................................................... 58 Commentary to Section 15.7.6.1 ............................................................................................................ 59 Additional Reference for Chapter 15 Commentary ................................................................................ 59 Chapter 16, Seismic Response History Procedures Section 16.1.3.2, Three-dimensional Analysis ....................................................................................... 61 Commentary to Section 16.1.3.2 ............................................................................................................ 61 Additional References for Chapter 16 Commentary .............................................................................. 62 Chapter 18, Seismic Design Requirements for Structures with Damping Systems Section 18.3.1, Nonlinear Response History Procedure ......................................................................... 63 Section 18.3.2, Nonlinear Static Procedure ............................................................................................ 63 Chapter 19, Soil-Structure Interaction for Seismic Design Table 19.2-1, Values of G/Go and Vs/Vso ............................................................................................... 65 Figure 19.2-1, Foundation Damping Factor ........................................................................................... 65 Chapter 21, Site-Specific Ground Motion Procedures for Seismic Design Section 21.2, Ground Motion Hazard Analysis ...................................................................................... 67 Commentary to Section 21.2 .................................................................................................................. 67 Additional References for Chapter 21 Commentary .............................................................................. 70 Chapter 22, Seismic Ground Motion and Long-period Transition Maps Chapter 22 ............................................................................................................................................. 73 Commentary for New Chapter 22 .......................................................................................................... 92 Chapter 23, Seismic Design Reference Documents Section 23.1, Consensus Standards and Other Reference Documents ................................................... 95 New Chapter 23, Vertical Ground Motions for Seismic Design Chapter 23 ............................................................................................................................................. 97 Commentary for New Chapter 23 .......................................................................................................... 98 2009 NEHRP RECOMMENDED SEISMIC PROVISIONS FOR NEW BUILDINGS AND OTHER STRUCTURES: PART 2, COMMENTARY TO ASCE/SEI 7-05 ...................................................................... 101 Commentary to Chapter 11, Seismic Design Criteria C11.1 General .................................................................................................................................................... 103 C11.1.1 Purpose ........................................................................................................................................ 103 C11.1.2 Scope ........................................................................................................................................... 103 C11.1.4 Alternate Materials and Alternate Means and Methods of Construction ...................................... 103 C11.4 Seismic Ground Motion Values .................................................................................................................. 104 C11.4.1 Mapped Acceleration Parameters ................................................................................................. 104 C11.4.3 and C11.4.4 Site Coefficient and Adjusted Acceleration Parameters ........................................... 104 C11.4.5 Design Response Spectrum .......................................................................................................... 105 C11.4.7 Site-Specific Ground Motion Procedures ..................................................................................... 105 C11.5 Importance Factor and Occupancy Category .............................................................................................. 105 C11.5.1 Importance Factor ......................................................................................................................... 106 C11.5.2 Protected Access for Category IV Structures ............................................................................... 106 C11.6 Seismic Design Categories .......................................................................................................................... 107 C11.7 Design Requirements for Seismic Design Category A ................................................................................ 108 C11.8 Geologic Hazards and Geotechnical Investigation ...................................................................................... 109 C11.8.1 Site Limitation for Seismic Design Categories E And F ............................................................... 109 C11.8.3 Additional Geotechnical Investigation Report Requirements for123 Seismic Design Categories D through F ...................................................................................... 109 References .................................................................................................................................................... 109 Commentary to Chapter 12, Seismic Design Requirements for Building Structures C12.1 Structural Design Basis ............................................................................................................................... 111 C12.1.1 Basic Requirements..................................................................................................................... 111 C12.1.2 Member Design, Connection Design, and Deformation Limit ..................................................... 113 C12.1.3 Continuous Load Path and Interconnection .................................................................................. 114 C12.1.4 Connection to Supports ................................................................................................................. 114 C12.1.5 Foundation Design ........................................................................................................................ 114 C12.1.6 Material Design and Detailing Requirements ............................................................................... 114 C12.2 Structural System Selection ......................................................................................................................... 114 C12.2.1 Selection and Limitations ............................................................................................................. 114 C12.2.2 Combinations of Framing Systems in Different Directions .......................................................... 116 C12.2.3 Combinations of Framing Systems in the Same Direction ........................................................... 116 C12.2.4 Combination of Framing Detailing Requirements ........................................................................ 116 C12.2.5 System Specific Requirements ..................................................................................................... 116 C12.3 Diaphragm Flexibility, Configuration Irregularities, and Redundancy ....................................................... 117 C12.3.1 Diaphragm Flexibility ................................................................................................................... 117 C12.3.2 Irregular and Regular Classification ............................................................................................. 118 C12.3.3 Limitations and Additional Requirements for Systems with Structural Irregularities .................. 119 C12.3.4 Redundancy ................................................................................................................................. 121 C12.4 Seismic Load Effects and Combinations ..................................................................................................... 123 C12.4.1 Applicability ................................................................................................................................ 123 C12.4.2 Seismic Load Effect ...................................................................................................................... 123 C12.4.3 Seismic Load Effect Including Overstrength Factor ..................................................................... 123 C12.4.4 Minimum Upward Force For Horizontal Cantilevers for Seismic Design Categories D Through F .................................................................................... 123 C12.5 Direction of Loading .................................................................................................................................. 124 C12.6 Analysis Selection Procedure ...................................................................................................................... 124 C12.7 Modeling Criteria ....................................................................................................................................... 124 C12.7.1 Foundation Modeling .................................................................................................................... 124 C12.7.2 Effective Seismic Weight .............................................................................................................. 125 C12.7.3 Structural Modeling ...................................................................................................................... 125 C12.7.4 Interaction Effects ......................................................................................................................... 126 C12.8 Equivalent Lateral Force Procedure ............................................................................................................ 126 C12.8.1 Seismic Base Shear ....................................................................................................................... 126 C12.8.2 Period Determination .................................................................................................................... 127 C12.8.3 Vertical Distribution of Seismic Forces ........................................................................................ 128 C12.8.4 Horizontal Distribution of Forces ................................................................................................. 129 C12.8.6 Story Drift Determination ............................................................................................................. 130 C12.8.7 P-delta Effects ............................................................................................................................... 132 C12.9 Modal Response Spectrum Analysis ........................................................................................................... 133 C12.9.1 Number of Modes ......................................................................................................................... 133 C12.9.2 Modal Response Parameters ......................................................................................................... 133 C12.9.3 Combined Response Parameters ................................................................................................... 133 C12.9.4 Scaling Design Values of Combined Response ............................................................................ 134 C12.9.5 Horizontal Shear Distribution ....................................................................................................... 134 C12.9.6 P-delta Effects ............................................................................................................................... 134 C12.10 Diaphragms, Chords, and Collectors ........................................................................................................... 134 C12.10.1 Diaphragm Design ...................................................................................................................... 134 C12.11 Structural Walls and Their Anchorage ........................................................................................................ 135 C12.11.1 Design for Out-of- Plane Forces ................................................................................................. 135 C12.11.2 Anchorage of Concrete or Masonry Structural Walls ................................................................. 136 C12.12 Drift and Deformation ................................................................................................................................ 137 C12.12.3 Building Separation ..................................................................................................................... 138 C12.12.4 Deformation Compatibility for Seismic Design Categories D Through F .................................. 139 C12.13 Foundation Design ...................................................................................................................................... 139 C12.13.3 Foundation Load-Deformation Characteristics ........................................................................... 139 C12.13.4 Reduction of Foundation Overturning ........................................................................................ 140 C12.13.5 Requirements for Structures Assigned to Seismic Design Category C ....................................... 140 C12.13.6 Requirements for Structures Assigned to Seismic Design Categories D Through F .................. 141 C12.14 Simplified Alternative Structural Design Criteria for Simple Bearing Wall or Building Frame Systems ............................................................................................................... 142 C12.14.1 General ....................................................................................................................................... 142 C12.14.3 Seismic Load Effects and Combinations..................................................................................... 142 C12.14.7 Design and Detailing Requirements ............................................................................................ 143 C12.14.8 Simplified Lateral Force Analysis Procedure ............................................................................. 143 References .................................................................................................................................................... 143 Commentary to Chapter 13, Seismic Design Requirements for Nonstructural Components C13.1 General .................................................................................................................................................... 145 C13.1.1 Scope ............................................................................................................................................ 147 C13.1.2 Seismic Design Category .............................................................................................................. 148 C13.1.3 Component Importance Factor ...................................................................................................... 148 C13.1.4 Exemptions .................................................................................................................................. 149 C13.1.5 Applicability of Nonstructural Component and Requirements ..................................................... 150 C13.1.6 Reference Documents ................................................................................................................... 150 C13.1.7 Reference Documents Using Allowable Stress Design ................................................................. 150 C13.2 General Design Requirements ..................................................................................................................... 151 C13.2.1 Applicable Requirements for Architectural, Mechanical, and Electrical Components, Supports, and Attachments ............................................................................................................ 151 C13.2.2 Special Certification Requirements For Designated Seismic Systems ......................................... 151 C13.2.3 Consequential Damage ................................................................................................................. 151 C13.2.4 Flexibility ..................................................................................................................................... 152 C13.2.5 Testing Alternative for Seismic Capacity Determination ............................................................. 152 C13.2.6 Experience Data Alternative for Seismic Capacity Determination ............................................... 153 C13.2.7 Construction Documents ............................................................................................................... 153 C13.3 Seismic Demands on Nonstructural Components ....................................................................................... 154 C13.3.1 Seismic Design Force ................................................................................................................... 154 C13.3.2 Seismic Relative Displacements ................................................................................................... 156 C13.4 Nonstructural Component Anchorage ......................................................................................................... 157 C13.4.2 Anchors in Concrete or Masonry .................................................................................................. 157 C13.4.3 Installation Conditions .................................................................................................................. 158 C13.4.4 Multiple Attachments ................................................................................................................... 158 C13.4.5 Power Actuated Fasteners ............................................................................................................. 158 C13.4.6 Friction Clips ............................................................................................................................... 158 C13.5 Architectural Components ........................................................................................................................... 159 C13.5.1 General ......................................................................................................................................... 159 C13.5.2 Forces and Displacements ............................................................................................................ 159 C13.5.3 Exterior Nonstructural Wall Elements and Connections .............................................................. 159 C13.5.5 Out-of-Plane Bending ................................................................................................................... 160 C13.5.6 Suspended Ceilings ....................................................................................................................... 160 C13.5.7 Access Floors ............................................................................................................................... 161 C13.5.8 Partitions ...................................................................................................................................... 161 C13.5.9 Glass in Glazed Curtain Walls, Glazed Storefronts, and Glazed Partitions .................................. 161 C13.6 Mechanical and Electrical Components ...................................................................................................... 162 C13.6.1 General ......................................................................................................................................... 163 C13.6.2 Component Period ........................................................................................................................ 163 C13.6.3 Mechanical Components and C13.6.4 Electrical Components ..................................................... 164 C13.6.5 Component Supports .................................................................................................................... 164 C13.6.6 Utility and Service Lines .............................................................................................................. 164 C13.6.7 HVAC Ductwork .......................................................................................................................... 165 C13.6.8 Piping Systems............................................................................................................................. 165 C13.6.9 Boilers and Pressure Vessels ........................................................................................................ 165 C13.6.10 Elevator and Escalator Design Requirements ............................................................................. 165 C13.6.11 Other Mechanical and Electrical Components ............................................................................ 166 References .................................................................................................................................................... 166 Commentary to Chapter 14, Material Specific Seismic Design and Detailing Requirements C14.0 Scope .................................................................................................................................................... 169 C14.1 Steel .................................................................................................................................................... 169 C14.1.1 Reference Documents ................................................................................................................... 169 C14.1.2 Seismic Design Categories B and C ............................................................................................. 169 C14.1.3 Seismic Design Categories D Through F ...................................................................................... 169 C14.1.4 Cold-Formed Steel ........................................................................................................................ 169 C14.1.5 Prescriptive Framing .................................................................................................................... 170 C14.1.6 Steel Deck Diaphragms ................................................................................................................ 170 C14.1.7 Steel Cables ................................................................................................................................. 170 C14.1.8 Additional Detailing Requirements for Steel Piles in Seismic Design Categories D Through F ................................................................................................................ 170 C14.2 Concrete .................................................................................................................................................... 171 C14.3 Composite Steel and Concrete Structures ................................................................................................... 173 C14.3.1 Reference Documents ................................................................................................................... 173 C14.3.2 Metal-Cased Concrete Piles .......................................................................................................... 174 C14.4 Masonry 206 .............................................................................................................................................. 174 C14.4.2 R Factors ...................................................................................................................................... 174 C14.4.3 Classification of Shear Walls ........................................................................................................ 174 C14.4.6 Modifications to Chapter 2 of ACI 530/ASCE 5/TMS 402 .......................................................... 174 C14.4.7 Modifications to Chapter 3 of ACI 530/ASCE 5/TMS 402 .......................................................... 175 C14.4.8 Modifications to Chapter 6 of ACI 530/ASCE 5/TMS 402 .......................................................... 175 C14.4.9 Modifications to ACI 530.1/ASCE 6/TMS 602 ............................................................................ 175 C14.5 Wood .................................................................................................................................................... 175 C14.5.1 Reference Documents ................................................................................................................... 175 C14.5.2 Framing ........................................................................................................................................ 175 References .................................................................................................................................................... 176 Commentary to Chapter 15, Seismic Design Requirements for Nonbuilding Structures C15.1.1 Nonbuilding Structures ................................................................................................................. 179 C15.1.2 Design .......................................................................................................................................... 179 C15.1.3 Structural Analysis Procedure Selection ....................................................................................... 180 C15.2 Reference Documents ................................................................................................................................ 185 C15.3 Nonbuilding Structures Supported by Other Structures .............................................................................. 186 C15.3.1 Less Than 25 Percent of Combined Weight Condition ................................................................. 186 C15.3.2 Greater Than or Equal to 25 Percent Combined Weight Condition .............................................. 186 C15.4 Structural Design Requirements .................................................................................................................. 187 C15.4.1 Design Basis ................................................................................................................................. 187 C15.4.2 Rigid Nonbuilding Structures ....................................................................................................... 188 C15.4.3 Loads ............................................................................................................................................ 188 C15.4.4 Fundamental Period ...................................................................................................................... 188 C15.4.8 Site-Specific Response Spectra ..................................................................................................... 188 C15.5 Nonbuilding Structures Similar to Buildings ............................................................................................... 188 C15.5.1 General ......................................................................................................................................... 188 C15.5.2 Pipe Racks .................................................................................................................................... 188 C15.5.3 Steel Storage Racks ....................................................................................................................... 188 C15.5.4 Electrical Power Generating Facilities .......................................................................................... 189 C15.5.5 Structural Towers for Tanks And Vessels ..................................................................................... 189 C15.5.6 Piers and Wharves ......................................................................................................................... 189 C15.6 General Requirements for Nonbuilding Structures Not Similar To Buildings ............................................ 190 C15.6.1 Earth-Retaining Structures ............................................................................................................ 190 C15.6.2 Stacks and Chimneys .................................................................................................................... 190 C15.6.4 Special Hydraulic Structures ......................................................................................................... 190 C15.6.5 Secondary Containment Systems .................................................................................................. 191 C15.6.6 Telecommunications Towers ........................................................................................................ 191 C15.7 Tanks and Vessels ...................................................................................................................................... 191 C15.7.1 General ......................................................................................................................................... 191 C15.7.2 Design Basis ................................................................................................................................. 192 C15.7.3 Strength and Ductility ................................................................................................................... 193 C15.7.4 Flexibility of Piping Attachments ................................................................................................. 193 C15.7.5 Anchorage .................................................................................................................................... 193 C15.7.6 Ground-Supported Storage Tanks for Liquids .............................................................................. 193 C15.7.7 Water Storage and Water Treatment Tanks and Vessels .............................................................. 195 C15.7.8 Petrochemical and Industrial Tanks and Vessels Storing Liquids ................................................ 195 C15.7.9 Ground-Supported Storage Tanks for Granular Materials ............................................................ 196 C15.7.10 Elevated Tanks and Vessels for Liquids and Granular Materials ............................................... 197 C15.7.11 Boilers and Pressure Vessels ....................................................................................................... 197 C15.7.12 Liquid and Gas Spheres .............................................................................................................. 197 C15.7.13 Refrigerated Gas Liquid Storage Tanks and Vessels .................................................................. 197 C15.7.14 Horizontal, Saddle Supported Vessels for Liquid or Vapor Storage ........................................... 198 References .................................................................................................................................................... 199 Commentary to Chapter 16, Seismic Response History Procedures C16.1 Linear Response History Procedure ............................................................................................................ 201 C16.1.1 Analysis Requirements ................................................................................................................. 201 C16.1.2 Modeling ...................................................................................................................................... 201 C16.1.3 Ground Motion ............................................................................................................................. 202 C16.1.4 Response Parameters .................................................................................................................... 203 C16.2 Nonlinear Response History Procedure ....................................................................................................... 204 C16.2.1 Analysis Requirements ................................................................................................................. 204 C16.2.2 Modeling ...................................................................................................................................... 204 C16.2.3 Ground Motion and Other Loading .............................................................................................. 205 C16.2.4 Response Parameters..................................................................................................................... 205 C16.2.5 Design Review .............................................................................................................................. 206 References .................................................................................................................................................... 206 Commentary to Chapter 17, Seismic Design Requirements for Seismically Isolated Structures C17.1 General .................................................................................................................................................... 207 C17.1.1 Variations in Material Properties .................................................................................................. 208 C17.2 General Design Requirements ..................................................................................................................... 208 C17.2.4 Isolation System........................................................................................................................... 209 C17.2.5 Structural System .......................................................................................................................... 209 C17.2.6 Elements of Structures and Nonstructural Components ............................................................... 209 C17.3 Ground Motion for Isolated Structures ........................................................................................................ 209 C17.3.1 Design Spectra .............................................................................................................................. 209 C17.3.2 Ground Motion Histories .............................................................................................................. 210 C17.4 Analysis Procedure Selection ...................................................................................................................... 210 C17.5 Equivalent Lateral Force Procedure ............................................................................................................ 211 C17.5.3 Minimum Lateral Displacements .................................................................................................. 211 C17.5.4 Minimum Lateral Forces .............................................................................................................. 212 C17.5.5 Vertical Distribution of Forces ..................................................................................................... 213 C17.5.6 Drift Limits .................................................................................................................................. 213 C17.6 Dynamic Analysis Procedures ..................................................................................................................... 214 C17.7 Design Review ........................................................................................................................................... 214 C17.8 Testing .................................................................................................................................................... 214 C17.8.5 Design Properties of the Isolation System .................................................................................... 215 References .................................................................................................................................................... 216 Commentary to Chapter 18, Seismic Design Requirements for Structures with Damping Systems C18.1 General .................................................................................................................................................... 217 C18.2 General Design Requirements ..................................................................................................................... 217 C18.2.2 System Requirements ................................................................................................................... 217 C18.2.4 Procedure Selection ...................................................................................................................... 217 C18.3 Nonlinear Procedures ................................................................................................................................. 219 C18.4 Response Spectrum Procedures and C18.5 Equivalent Lateral Force Procedure ........................................ 219 C18.6 Damped Response Modification ................................................................................................................. 221 C18.6.1 Damping Coefficient.................................................................................................................... 221 C18.6.2 Effective Damping ........................................................................................................................ 222 C18.7 Seismic Load Conditions and Acceptance Criteria ..................................................................................... 222 References .................................................................................................................................................... 222 Commentary to Chapter 19, Soil Structure Interaction for Seismic Design C19.1 General .................................................................................................................................................... 225 C19.2 Equivalent Lateral Load Procedure ............................................................................................................. 226 C19.2.1 Base Shear ................................................................................................................................... 226 C19.2.2 Vertical Distribution of Seismic Forces ...................................................................................................... 228 C19.2.3 Other Effects .............................................................................................................................................. 229 C19.3 Modal Analysis Procedure ........................................................................................................................... 229 References .................................................................................................................................................... 229 Commentary to Chapter 20, Site Classification Procedure for Seismic Design C20.1 Site Classification ....................................................................................................................................... 231 C20.3 Site Class Definitions ................................................................................................................................. 231 C20.3.1 Site Class F ................................................................................................................................... 231 C20.3.2 through C20.3.5 ............................................................................................................................. 231 C20.4 Definitions of Site Class Parameters ........................................................................................................... 231 Commentary to Chapter 21, Site-Specific Ground Motion Procedures for Seismic Design General .................................................................................................................................................... 233 C21.1 Site Response Analysis ............................................................................................................................... 233 C21.1.1 Base Ground Motions ................................................................................................................... 233 C21.1.2 Site Condition Modeling ............................................................................................................... 234 C21.1.3 Site Response Analysis and Computed Results ............................................................................ 234 C21.2 Ground Motion Hazard Analysis ................................................................................................................. 234 C21.2.1 Probabilistic MCE ......................................................................................................................... 235 C21.2.2 Deterministic MCE ....................................................................................................................... 235 C21.3 Design Response Spectrum ......................................................................................................................... 235 C21.4 Design Acceleration Parameters .................................................................................................................. 235 References .................................................................................................................................................... 235 Commentary to Chapter 22, Seismic Ground Motion and Long-Period Transition Maps Seismic Ground Motion Maps .................................................................................................................................. 237 Long-Period Transition Maps ................................................................................................................................... 237 References .................................................................................................................................................... 238 2009 NEHRP RECOMMENDED SEISMIC PROVISIONS FOR NEW BUILDINGS AND OTHER STRUCTURES: PART 3, RESOURCE PAPERS (RP) ON SPECIAL TOPICS IN SEISMIC DESIGN ....... 239 RP 1 Alternate Materials, Design, and Methods of Construction ........................................................................ 241 RP 2 Nonlinear Static Procedure ......................................................................................................................... 243 RP 3 Seismic Response-History Analysis ........................................................................................................... 247 RP 4 Foundation Geotechnical Ultimate Strength Design of Foundations and Foundation Load-Deformation Modeling .................................................................................................... 251 RP 5 Alternative Provisions for the Design of Piping Systems ............................................................................ 259 RP 6 Other Nonbuilding Structures ..................................................................................................................... 263 RP 7 Special Requirements for Seismic Design of Structural Glued Laminated Timber (Gluam) Arch Members and Their Connections in Three-Hinge Arch Systems ....................................................... 267 RP 8 Appropriate Seismic Load Combinations for Base Plates, Anchorage, and Foundations .......................... 275 RP 9 Seismic Design Using Target Drift, Ductility, and Plastic Mechanisms as Performance Criteria .................................................................................................................... 289 RP 10 Seismic Design Methodology for Precast Concrete Floor Diaphragms ...................................................... 311 RP 11 Shear Wall Load-Deflection Parameters and Performance Expectations .................................................... 333 RP 12 Evaluation of Geologic Hazards and Determination of Seismic Lateral Earth Pressures ............................ 341 RP 13 Light-Frame Wall Systems with Wood Structural Panel Sheathing ............................................................ 365 APPENDIX, Participants in the BSSC 2009 Provisions Update Project ............................................................. 373 2009 NEHRP RECOMMENDED SEISMIC PROVISIONS FOR NEW BUILDINGS AND OTHER STRUCTURES: PART 1, PROVISIONS Work on this 2009 edition of the NEHRP (National Earthquake Hazards Reduction Program) Recommended Seismic Provisions for New Buildings and Other Structures began in September 2004 when the National Institute of Building Sciences, the parent organization of the Building Seismic Safety Council (BSSC), entered into a contract with Federal Emergency Management Agency (FEMA) for initiation of the 2009 Provisions update effort. During 2005, the BSSC member organization representatives and alternate representatives and the BSSC Board of Direction were asked to identify individuals to serve on the 2009 Provisions Update Committee (PUC) and its Technical Subcommittees (TSs) and to suggest topics for concentrated study by ad hoc Issue Teams. The 2009 PUC and its eight Technical Subcommittees (TS) then were established to address document composition and management; design criteria and analysis and advanced technologies; mapping, foundations, and geotechnical considerations; concrete structures; masonry structures; steel and composite steel and concrete structures; wood structures; nonstructural components and nonbuilding structures. Three Issue Teams (ITs) also were established to focus on performance criteria, design parameters, and foundation design requirements. Further, given ongoing work by the U.S. Geological Survey (USGS) to update its seismic hazard maps, a Seismic Design Procedures Review Group (SDPRG) was established to consider the emerging maps and to re-examine the existing design maps and procedures that were introduced in the 1997 edition of the Provisions and that remained essentially unchanged for the 2000 and 2003 editions of the Provisions. Work already done and decisions made prior to initiation of the 2009 Provisions update project recognized that the codes and standards arena has changed over the past decade and that those changes called for a refocusing of the Provisions on exploration of new technologies and procedures and less consideration of format and editorial changes. To this end, the initial efforts of the 2009 PUC and its TSs focused on adoption of the national load standard, Minimum Design Loads for Building and Other Structures, ASCE/SEI 7-05 (including Supplements No. 1 and No. 2), as the primary reference standard of the Provisions and the identification of parts of the 2003 Provisions that should be maintained as modifications to the standard or otherwise revised to reflect new knowledge and experience data. The result of this effort was a vote by the BSSC member organizations to adopt ASCE/SEI 7-05 by reference and for it to serve as the base document for the update cycle. Three modifications to standard, originally appendices to various chapters of the 2003 Provisions, had been deemed needed by the PUC and TSs and were approved as part of this vote by the membership for inclusion in the 2009 Provisions. As the update cycle progressed, additional modifications to the standard were prepared and voted on by the membership in three separate ballots. All these modifications appear in Part 1 of this document and, together with ASCE/SEI 7-05 and the references cited therein, constitute the 2009 Provisions. (A summary of the results of the member organization ballots and comment resolution process is available from the BSSC upon written request and will be posted on the BSSC website at www.bssconline.org.) A major effort also was made to rewrite the commentary to the Provisions. Until now, the commentary was published in a separate volume and tended to explain the development of the requirements. For 2009, the commentary appears in Part 2 of this Provisions volume and explains how to apply the Provisions requirements as articulated in ASCE/SEI 7-05 and the references cited therein. (Note that the Part 1 modifications to the standard are accompanied by appropriate commentary text included in Part 1.) Part 3 of this Provisions volume introduces new procedures or provisions not currently contained in the referenced standards for consideration and experimental use by the design community, researchers, and standards- and codedevelopment organizations and feedback from these users is encouraged. Part 3 also presents individual summaries of ongoing committee work that awaits additional research before being submitted to the BSSC membership for consensus approval and provides useful guidance on the application of Part 1 requirements, either as a discussion of an overall approach or as a detailed procedure. 1.1 INTENT The NEHRP Recommended Seismic Provisions for New Buildings and Other Structures presents the minimum recommended requirements necessary for the design and construction of new buildings and other structures to resist earthquake ground motions throughout the United States. The intent of these provisions is to provide reasonable assurance of seismic performance that will: 1. Avoid serious injury and life loss, 2. Avoid loss of function in critical facilities, and 3. Minimize structural and nonstructural repair costs where practical to do so. These objectives are addressed by seeking to avoid structural collapse in very rare, extreme ground shaking and by seeking to provide reasonable control of damage to structural and nonstructural systems that could lead to injury and economic or functionality losses for more moderate and frequent ground shaking. These design requirements include minimum lateral strength and stiffness for structural systems and guidance for anchoring, bracing, and accommodation of structural drift for nonstructural systems. Occupancy Category III or IV structures intended to provide enhanced safety and functionality are required to have more strength than Occupancy Category I or II structures in an effort to reduce damage to the structural system. Allowable drifts are reduced to control damage to building components connected to multiple floor levels. Nonstructural system performance is enhanced by strengthening the anchorage and bracing requirements, and important equipment must be shown to be functional after being shaken. The degree to which these goals can be achieved depends on a number of factors including structural framing type, building configuration, materials, as-built details, and overall quality of design. In addition, large uncertainties as to the intensity and duration of shaking and the possibility of unfavorable response of a small subset of buildings or other structures may prevent full realization of the intent. 1.2 REFERENCE DOCUMENT Design for seismic resistance of structural elements including foundation elements and nonstructural components shall conform to the requirements of ASCE/SEI 7-05, Minimum Design Loads for Buildings and Other Structures, including Supplements No. 1 and No. 2 (referred to hereinafter as ASCE/SEI 7-05), as modified herein.1 COMMENTARY TO SECTIONS 1.1 AND 1.2 1 Supplement No. 2 of the standard is available for download at http://content.seinstitute.org/files/pdf/SupplementNo2ofthe2005Editionof ASCE7.pdf. 2 The derivation of MCE ground motion was described in detail in Commentary Appendix A of the 2003 NEHRP Recommended Provisions (FEMA 450-2), and this appendix, “Development of Maximum Considered Earthquake Ground Motion Maps Figures 3.3-1 through 3.3-14,” can be downloaded from http://www.nibs.org/index.php/bssc/publications/fema450nehrp2003/. The primary intent of the NEHRP Recommended Seismic Provisions for New Buildings and Other Structures is to prevent, for typical buildings and structures, serious injury and life loss caused by damage from earthquake ground shaking. Most earthquake injuries and deaths are caused by structural collapse; therefore, the major thrust of the Provisions is to prevent collapse for very rare, intense ground motion, termed the maximum considered earthquake (MCE) ground motion.2 The intent remains the same in the 2009 Provisions; however, the prevention of collapse is redefined in terms of risk-targeted maximum considered earthquake (MCER) ground motions. This change is explained fully in the commentary to the Part 1 modification to ASCE/SEI 7-05 Section 11.2. Falling exterior walls and cladding and falling interior ceilings, light fixtures, pipes, equipment, and other nonstructural components also cause deaths and injuries. The Provisions minimizes this risk using requirements for anchoring and bracing nonstructural components, although this level of protection generally is aimed at ground motions less severe than the MCER ground motion. This anchoring and bracing of nonstructural systems coupled with reasonable limitations on differential movement between floors (i.e., story drift limits) also serve to control damage that may be costly to repair or that would result in lengthy building closures, particularly for moderate shaking levels. Stricter story drift limits can further limit damage to components connected to more than one floor (e.g., walls, cladding and stairways) but, at the same time, can create higher acceleration levels in the building that could increase damage to nonstructural components braced or anchored to a single floor (e.g., ceilings, light fixtures, and pipes). Achieving an optimum balance between the cost and performance of the structural system and the effect of structural stiffness on performance of the nonstructural systems is impossible using the prescriptive rules of a building code, particularly given the variety of structural systems used in the United States. Buildings deemed to have higher importance due to hazardous contents or critical occupancy are assigned to higher Occupancy Categories (see ASCE/SEI 7-05 Table 1-1). The damage level in these buildings is intended to be reduced by decreasing nonlinear demand using an importance factor, I, to reduce the response modification coefficient, R. The resulting increased strength will reduce structural damage, or increase reliability of acceptable performance, for a given level of shaking. Some authorities having jurisdiction subject the design and construction of such buildings to a higher level of scrutiny. The performance of critical occupancy structures in past earthquakes indicates that the increase in the importance factor controls structural damage in moderate shaking. In strong shaking associated with the design level of two-thirds the maximum considered earthquake or higher, the values of I have not been well tested for their effect on either functionality for critical buildings or increased reliability of life safety protection for high occupancy buildings. The importance factor also increases the design anchorage and bracing load for nonstructural systems, which should increase the reliability of their staying in place and, thus, remaining undamaged. In addition, certain critical equipment must remain operable after strong shaking. Experience data show that some nonstructural components will remain functional if they stay in position, but other components will require testing to show that they will function following strong shaking. The emphasis to date has been on the seismic qualification of individual components. However, the nonstructural systems of many buildings are, in reality, complex networks that can be shut down by a single failure. For example, a break in a pressurized pipe can flood part or all of a building forcing it to close, and failure of the anchorage (or internal workings) of a battery, day tank, fuel lines, muffler, or main engine can shut down an emergency generator. Therefore, the special regulations for seismic protection of nonstructural systems represent a rational approach to achieving performance appropriate for the various occupancies, but experience data to confirm their adequacy are lacking. When the hazard definition for design was changed from motion with a 2 percent chance of exceedance in 50 years to the 1 percent chance of collapse in 50 years, the primary intended performance was retained. The design basis ground motion is still two-thirds of the risk-targeted maximum considered earthquake (MCER) ground motion. The increase in the importance factor is intended to ensure a lower probability of collapse for the performance of higher occupancy and critical buildings. The Provisions requirements are not intended to prevent damage due to earth slides (such as those that occurred in Anchorage, Alaska) or tsunami (such as occurred in Hilo, Hawaii, and the Indian Ocean). They provide only for required resistance to earthquake ground shaking without significant settlement, slides, subsidence, or faulting in the immediate vicinity of the structure. In most cases, practical engineering solutions are available to resist other potential earthquake hazards, but they must be developed on a case-by-case basis. Although the Provisions sets the minimum performance goals described in Section 1.1, earthquake performance of buildings and other structures is highly variable. The characteristics of the shaking itself are highly uncertain and even different sets of motions defined to qualify as maximum considered earthquake ground motions can result in significantly different responses. Additional uncertainty is created by the wide variety of systems and configurations allowed under the regulations as well as by the various interpretations and implementation practices of individual designers. Thus, a small percentage of buildings designed to the requirements of the Provisions may not meet the performance intent when exposed to earthquake ground motions. The commentary the Tentative Provisions for the Development of Seismic Regulations for Buildings (Applied Technology Council, 1978), upon which the first edition of the NEHRP Recommended Provisions (1985) was based, suggested a less than 1 percent chance of collapse in a 50-year period for a building designed using the tentative requirements. More recent studies (e.g., Quantification of Building Seismic Performance Factors, FEMA P-695, 2009) suggest a 10 percent chance of collapse with shaking at the maximum considered earthquake level, which is roughly equivalent to the 1978 estimations. 1.3 MODIFICATIONS TO ASCE/SEI 7-05 With only a few exceptions (such as the changes to Table 12.2-1 shown in underline and strikeout), modifications are presented as replacements for existing sections of ASCE/SEI 7-05 or as new sections to be added to the standard. Commentary, if any, to the modifications that appear in the remainder of this part of the 2009 Provisions is presented at the end of each chapter of modifications. Commentary to the seismic chapters (Chapters 11 through 22) of the unmodified reference document itself, ASCE/SEI 7-05, is presented in Part 2 of the 2009 Provisions. Modifications to Chapter 11, Seismic Design Criteria Replace with the following: SECTION 11.1.2, SCOPE 11.1.2 Scope. Every structure, and portion thereof, including nonstructural components, shall be designed and constructed to resist the effects of earthquake motions as prescribed by the seismic requirements of this standard. Certain nonbuilding structures, as described in Chapter 15, are also within the scope and shall be designed and constructed in accordance with the requirements for Chapter 15. Requirements concerning alterations, additions, and change of use are set forth in Appendix 11B. Existing structures and alterations to existing structures need only comply with the seismic requirements of this standard where required by Appendix 11B. The following structures are exempt from the seismic requirements of this standard: 1. Detached one- and two-family dwellings that are located where the mapped, short period, spectral response acceleration parameter, SS, is less than 0.4 or where the Seismic Design Category determined in accordance with Section 11.6 is A, B or C. 2. Dwellings of wood-frame construction satisfying the limitations of and constructed in accordance with the International Residential Code. 3. Buildings of wood-frame construction satisfying the limitations of and constructed in accordance with Section 2308 of the International Building Code. 4. Agricultural storage structures that are intended only for incidental human occupancy. 5. Structures that require special consideration of their response characteristics and environment that are not addressed in Chapter 15 and for which other regulations provide seismic criteria, such as vehicular bridges, electrical transmission towers, hydraulic structures, buried utility lines and their appurtenances, and nuclear reactors. SECTION 11.2, DEFINITIONS Change the definition for “Maximum Considered Earthquake (MCE) Ground Motion” to: RISK-TARGETED MAXIMUM CONSIDERED EARTHQUAKE (MCER) GROUND MOTION: The most severe earthquake effects considered by this standard as defined in Section 11.4. Add the following new definition: MAXIMUM CONSIDERED EARTHQUAKE GEOMETRIC MEAN PEAK GROUND ACCELERATION (PGAM): The most severe earthquake effects considered for liquefaction as defined in Section 11.8. SECTION 11.3, NOTATION Add the following: CR = risk coefficient; see Section 21.2.1.1 CRS = mapped value of the risk coefficient at short periods as defined by Figure 22-3 CR1 = mapped value of the risk coefficient at a period of 1 second as defined by Figure 22-4 SSD = mapped deterministic, 5 percent damped, spectral response acceleration parameter at short periods as defined in Section 11.4.1 SSUH = mapped uniform-hazard, 5 percent damped, spectral response acceleration parameter at short periods as defined in Section 11.4.1 S1D = mapped deterministic, 5 percent damped, spectral response acceleration parameter at a period of 1 second as defined in Section 11.4.1 S1UH = mapped uniform-hazard, 5 percent damped, spectral response acceleration parameter at a period of 1 second as defined in Section 11.4.1 Revise the following to read as indicated: SS = 5 percent damped, spectral response acceleration parameter at short periods as defined in Section 11.4.3 S1 = spectral response acceleration parameter at a period of 1 second as defined in Section 11.4.3 SaM = the site-specific MCER spectral response acceleration at any period SMS = the MCER, 5 percent damped, spectral response acceleration parameter at short periods adjusted for target risk and site-class effects as defined in Section 11.4.3 SM1 = the MCER, 5 percent damped, spectral response acceleration parameter at a period of 1 second adjusted for target risk and site-class effects as defined in Section 11.4.3 SECTION 11.4, SEISMIC GROUND MOTION Replace with the following: 11.4 SEISMIC GROUND MOTION VALUES (11.4-1) (11.4-2) and the spectral response acceleration at a period of 1 second (S1), adjusted for the target risk of collapse, shall be determined as the lesser value of Equations 11.4-3 and 11.4-4: (11.4-3) (11.4-4) where SSD = mapped deterministic, 5 percent damped, spectral response acceleration parameter at short periods as defined in Section 11.4.1 SSUH = mapped uniform-hazard, 5 percent damped, spectral response acceleration parameter at short periods as defined in Section 11.4.1 CRS = mapped value of the risk coefficient at short periods as defined in Section 11.4.1 S1D = mapped deterministic, 5 percent damped, spectral response acceleration parameter at a period of 1 second as defined in Section 11.4.1 3 11.4.1 Mapped Acceleration Parameters and Risk Coefficients. The parameters SSUH , S1UH, SSD, and S1D shall be determined from the 0.2- and 1-second spectral response accelerations shown on Figures 22-1 and 22-2 and Figures 22-5 through 22-6, respectively, and the risk coefficients CRS and CR1 shall be determined from Figures 22-3 and 22-4, respectively. 11.4.2 Site Class. Based on the site soil properties, the site shall be classified as either Site Class A, B, C, D, E, or F in accordance with Chapter 20. Where the soil properties are not known in sufficient detail to determine the Site Class, Site Class D shall be used unless the authority having jurisdiction or geotechnical data determines Site Class E or F soils are present at the site. 11.4.3 Site Coefficients, Risk Coefficients, and Risk-targeted Maximum Considered Earthquake (MCER) Spectral Response Acceleration Parameters. The spectral response acceleration for short periods (SS), adjusted for the target risk of collapse, shall be determined as the lesser value of Equations 11.4-1 and 11.4-2: Equation SS = CRS SSUH Equation S SD S = S Equation 1 R1 1UH S = C S Equation 1 1D S = S 3 To utilize the U.S. Geological Survey’s seismic design map web application to obtain ground motion values, visit http://earthquake.usgs.gov/designmaps/usapp. Also see the USGS introduction to the web application included on the CD that accompanies this volume. S1UH = mapped uniform-hazard, 5 percent damped, spectral response acceleration parameter at a period of 1 second as defined in Section 11.4.1 CR1 = mapped value of the risk coefficient at a period of 1 second as defined in Section 11.4.1 The MCER spectral response acceleration for short periods (SMS) and at 1 second (SM1), adjusted for Site Class effects and the target risk of collapse, shall be determined by Equations 11.4-5 and 11.4-6, respectively. SMS = FaSs (11.4-5) SM1 = FvS1 (11.4-6) where site coefficients Fa and Fv are defined in Tables 11.4-1 and 11.4-2, respectively. When the simplified design procedure of Section 12.14 is used, the value Fa shall be determined in accordance with Section 12.14.8.1, and the values of Fv, S1, SMS, and SM1 need not be determined. Where S1 is less than or equal to 0.04 and SS is less than or equal to 0.15, the structure is permitted to be assigned to Seismic Design Category A and is only required to comply with Section 11.7. Table 11.4-1 Site Coefficient, Fa Site Class Spectral Response Acceleration Parameter at Short Period SS = 0.25 SS = 0.5 SS = 0.75 SS = 1.0 SS = 1.25 A 0.8 0.8 0.8 0.8 0.8 B 1.0 1.0 1.0 1.0 1.0 C 1.2 1.2 1.1 1.0 1.0 D 1.6 1.4 1.2 1.1 1.0 E 2.5 1.7 1.2 0.9 0.9 F See Section 11.4.7 Note: Use straight-line interpolation for intermediate values of SS. Table 11.4-2 Site Coefficient, Fv Site Class Spectral Response Acceleration Parameter at 1-second Period S1 = 0.1 S1 = 0.2 S1 = 0.3 S1 = 0.4 S1 = 0.5 A 0.8 0.8 0.8 0.8 0.8 B 1.0 1.0 1.0 1.0 1.0 C 1.7 1.6 1.5 1.4 1.3 D 2.4 2.0 1.8 1.6 1.5 E 3.5 3.2 2.8 2.4 2.4 F See Section 11.4.7 Note: Use straight-line interpolation for intermediate values of S1. 11.4.4 Design Spectral Acceleration Parameters. Design earthquake spectral response acceleration parameter at short periods, SDS, and at a 1-second period, SD1, shall be determined from Equations 11.4-7 and 11.4-8, respectively. Where the alternate simplified design procedure of Section 12.14 is used, the value of SDS shall be determined in accordance with Section 12.14.8.1, and the value of SD1 need not be determined: (11.4-7) (11.4-8) 11.4.5 Design Response Spectrum. Where a design response spectrum is required by this standard and site-specific ground motion procedures are not used, the design response spectrum curve shall be developed as indicated in Figure 11.4-1 and as follows: Equation 2 DS 3 MS S = S Equation 1 1 2 D 3 M S = S 1. For periods less than T0, the design spectral response acceleration, Sa, shall be taken as given by Equation 11.4-9: (11.4-9) Equation 0 0 4 0.6 a DS S S . T T . . = +.. . . 2. For periods greater than or equal to T0 and less than or equal to TS, the design spectral response acceleration, Sa, shall be taken equal to SDS. Figure 11.4-1 Showing design response spectrum. 1.0 0 1 0 1 Period, T (sec) Spectral Response Acceleration, Sa (g) SDS SD1 D1 a S S T = T L 2 1T S SD TL a · = T 0 T S 3. For periods greater than TS, and less than or equal to TL, the design spectral response acceleration, Sa, shall be taken as given by Equation 11.4-10: (11.4-10) Equation D1 a S S T = 4. For periods greater than TL, Sa shall be taken as given by Equation 11.4-11: (11.4-11) Equation Equation where SDS = the design spectral response acceleration parameter at short periods; SD1 = the design spectral response acceleration parameter at 1-second period; T = the fundamental period of the structure, seconds; T0 = 0.2 SD1/SDS; TS = SD1/SDS; and TL = long-period transition period (seconds) shown in Figure 22-7. 11.4.6 MCER Response Spectrum. Where a MCER response spectrum is required, it shall be determined by multiplying the design response spectrum by 1.5. 11.4.7 Site-Specific Ground Motion Procedures. The site-specific ground motion procedures set forth in Chapter 21 are permitted to be used to determine ground motions for any structure. A site response analysis shall be performed in accordance with Section 21.1 for structures on Site Class F sites, unless the exception to Section 20.3.1 is applicable. For seismically isolated structures and for structures with damping systems on sites with S1 greater than or equal to 0.6, a ground motion hazard analysis shall be performed in accordance with Section 21.2. Figure 11.4-1 Design response spectrum. SECTION 11.8, GEOLOGIC HAZARDS AND GEOTECHNICAL INVESTIGATION Replace with the following: 11.8 Geologic Hazards and Geotechnical Investigation 11.8.1 Site Limitation for Seismic Design Categories E and F. A structure assigned to Seismic Design Category E or F shall not be located where there is a known potential for an active fault to cause rupture of the ground surface at the structure. 11.8.2 Geotechnical Investigation Report Requirements for Seismic Design Categories C through F. A geotechnical investigation report shall be provided for a structure assigned to Seismic Design Category C, D, E, or F in accordance with this section. An investigation shall be conducted and a report shall be submitted that shall include an evaluation of the following potential geologic and seismic hazards: 1. Slope instability; 2. Liquefaction; 3. Total and differential settlement; and 4. Surface displacement due to faulting or seismic-induced lateral spreading or lateral flow. The report shall contain recommendations for appropriate foundation designs or other measures to mitigate the effects of the above hazards. EXCEPTION: Where deemed appropriate by the authority having jurisdiction, a site-specific geotechnical report is not required when prior evaluations of nearby sites with similar soil conditions provide sufficient direction relative to the proposed construction. 11.8.3 Additional Geotechnical Investigation Report Requirements for Seismic Design Categories D through F. The geotechnical investigation report for a structure assigned to Seismic Design Category D, E, or F shall include: 1. The determination of dynamic seismic lateral earth pressures on basement and retaining walls due to design earthquake ground motions. 2. The potential for liquefaction and soil strength loss evaluated for site peak ground acceleration, earthquake magnitude, and source characteristics consistent with the maximum considered earthquake geometric mean peak ground accelerations. Peak ground acceleration shall be determined based on either: (a) a site-specific study taking into account soil amplification effects as specified in Section 11.4.7 or (b) the peak ground acceleration, PGAM, from Equation 11.8-1: PGAM = FPGA PGA (11.8-1) where PGAM = maximum considered earthquake geometric mean peak ground acceleration adjusted for Site Class effects; PGA = mapped maximum considered earthquake geometric mean peak ground acceleration shown in Figures 22-8 through 22-11; and FPGA = site coefficient from Table 11.8-1. 3. Assessment of potential consequences of liquefaction and soil strength loss as computed in Item 2, including estimation of total and differential settlement, lateral soil movement, lateral soil loads on foundations, reduction in foundation soil-bearing capacity and lateral soil reaction, soil downdrag and reduction in axial and lateral soil reaction for pile foundations, increases in soil lateral pressures on retaining walls, and flotation of buried structures. 4. Discussion of mitigation measures such as selection of appropriate foundation type and depths, selection of appropriate structural systems to accommodate anticipated displacements and forces, ground stabilization, or any combination of these measures and how they shall be considered in the design of the structure. Table 11.8-1 Site Coefficient FPGA Site Class Mapped MCE Geometric Mean Peak Ground Acceleration, PGA PGA = 0.1 PGA = 0.2 PGA = 0.3 PGA = 0.4 PGA = 0.5 A 0.8 0.8 0.8 0.8 0.8 B 1.0 1.0 1.0 1.0 1.0 C 1.2 1.2 1.1 1.0 1.0 D 1.6 1.4 1.2 1.1 1.0 E 2.5 1.7 1.2 0.9 0.9 F See Section 11.4.7 Note: Use straight-line interpolation for intermediate values of PGA. Commentary to Chapter 11 Modifications COMMENTARY TO SECTION 11.1.2 C11.1.2 Scope. The scope statement establishes in general terms the applicability of the standard as a base of reference. Certain structures are exempt and need not comply. The reasons for each are described below. Note that it is not acceptable to use a combination of International Building Code (IBC) and International Residential Code (IRC) conventional construction provisions. Conventional requirements of either the IBC or the IRC can be combined with engineered design of elements in accordance with IBC engineered design requirements. Elements designed using the IBC engineered design requirements are not exempt from the seismic requirements of ASCE/SEI 7. Exemption 1 – Detached one- and two-family dwellings in Seismic Design Categories A, B, and C, along with those located where Ss < 0.4g, are exempt because they represent low seismic risks. Exemption 2 – This exemption recognizes that the wood-frame seismic design requirements of the International Residential Code (IRC) substantially meet the intent of conventional construction (wood-frame) provisions included in the NEHRP Recommended Seismic Provisions through the 2003 Edition. Exemption 3 – This exemption recognizes that wood-frame seismic design requirements of International Building Code (IBC) Section 2308 substantially meet the intent of conventional construction (wood-frame) provisions included in the NEHRP Recommended Seismic Provisions through the 2003 Edition. Exemption 4 – Agricultural storage structures generally are exempt from most code requirements because of the exceptionally low risk to life involved. Exemption 5 – Bridges, transmission towers, nuclear reactors, and other structures with special configurations and uses are not covered because regulations developed to apply to buildings and building-like structures do not adequately address their design and performance issues. The standard is not retroactive and usually applies to existing structures only where there is an addition, change of use, or alteration. Minimum acceptable seismic resistance of existing buildings is a policy issue normally set by the authority having jurisdiction. Appendix 11B of the standard contains rules of application for basic conditions. ASCE 31, Seismic Evaluation of Buildings, and ASCE 41, Seismic Rehabilitation of Existing Buildings, are available for technical guidance but do not contain policy recommendations. The International Code Council includes a chapter in the IBC to control the alteration, repair, addition, and change of occupancy of existing buildings and also maintains the International Existing Building Code (IEBC) and an associated commentary. COMMENTARY TO SECTION 11.2 C11.2 DEFINITIONS Renaming the maximum considered earthquake (MCE) ground motions as the risk-targeted maximum considered earthquake (MCER) ground motions is an editorial change recommended by the BSSC’s Provisions Update Committee and accepted by the BSSC’s Board. The MCER ground motions are based on the 2008 USGS seismic hazard maps and also incorporate three technical changes to ASCE/SEI 7-05: 1. Use of risk-targeted ground motions, 2. Use of maximum direction ground motions, and 3. Use of near-source 84th percentile ground motions. Reasons for each of the three technical changes are included in the commentary that accompanies the modifications to Chapter 21. COMMENTARY TO SECTIONS 11.4.3 AND 11.4.4 C11.4.3 Site Coefficients, Risk Coefficients, and Risk-targeted Maximum Considered Earthquake (MCER) Spectral Response Acceleration Parameters. The following illustrates the process of developing MCER response spectral accelerations using the formulas and maps of Section 11.4.3 and Chapter 22, respectively, and provides a summary of design ground motions for 34 city sites in regions of the United States of greatest seismic risk. Additional information and references explaining the differences from the MCE ground motions in ASCE/SEI 7-05 are included in the commentary to Chapter 21. Illustration of the Development of MCER Spectral Response Acceleration Using Section 11.4.3 Equations and Chapter 22 Uniform-Hazard, Risk Coefficient, and Deterministic Maps. The formulas of Section 11.4.3 (and the associated uniform-hazard, risk coefficient, and deterministic maps of Chapter 22) are intended to add transparency to the development of MCER ground motions. The development of MCER ground motions is explained in Section 21.2 and its commentary as part of the site-specific ground motion procedures for seismic design. As will be illustrated, the formulas (and maps) add transparency by emulating the site-specific procedure. A cost of this transparency is the added complexity of more formulas (and maps). However, a USGS website similar to the USGS Java ground motion parameter calculator automates use of the proposed formulas (and maps): http://earthquake.usgs.gov/designmaps/usapp. The three steps that the website implements are as follows: Step 1 – Adjust uniform-hazard ground motions (Site Class B) for target risk of collapse As illustrated in the top row of Figure C11.4-1, the first step is to obtain the mapped uniform-hazard (2 percent-in-50-years probability of exceedance) spectral response acceleration for short periods (SSUH) from Figure 22-1 and for a period of 1 second (S1UH) from Figure 22-2, and then to multiply these values by the corresponding mapped risk coefficients (CRS and CR1) from Figures 22-3 and 22-4, respectively. This step is expressed in Equations 11.4-1 for the short periods and 11.4-3 for the 1-second period and is consistent with Section 21.2.1 of the site-specific procedure in Chapter 21. The resulting spectral response accelerations, CRS, SSUH and CR1S1UH, are referred to as probabilistic ground motions. Figure C11.4-1 illustrates this for the 1-second period only using small maps of the conterminous United States that depict S1UH, CR1, and CR1S1UH. The reasons for using 2 percent-in-50-years (uniform-hazard) spectral response accelerations, which were the basis for the probabilistic portions of the MCE ground motion maps in ASCE/SEI 7-05, are explained in the commentary of the 2003 NEHRP Recommended Provisions. As explained below in the Chapter 21 commentary, the uniform-hazard maps (Figures 22-1 and 22-2) represent the spectral response acceleration in the maximum direction, which are larger than the geometric mean spectral response acceleration maps developed by the USGS by factors of 1.1 for the short periods and 1.3 for the 1- second period. The risk coefficients adjust these uniform-hazard (2 percent-in-50-years) spectral response accelerations to achieve building designs with 1 percent probability of collapse in 50 years (i.e., uniform risk), as explained below in the Chapter 21 commentary. Step 2 – Take minimum of probabilistic and deterministic ground motions (Site Class B) As illustrated in the middle row of Figure C11.4-1, the second step in the development of MCER ground motions is to obtain the mapped deterministic spectral response acceleration for short periods (SSD) from Figure 22-5 and for a period of 1 second (S1D) from Figure 22-6, and then to take the minimum of each of these values (expressed in Equations 11.4-2 and 11.4-4, respectively) and the corresponding value resulting from Step 1 (i.e., those expressed in Equations 11.4-1 and 11.4-3, respectively). This step is consistent with Sections 21.2.2 (“Deterministic Ground Motions”) and 21.2.3 (“Site Specific MCER”) of the site-specific procedure in Chapter 21. The resulting spectral response accelerations are denoted SS for the short periods and S1 for the 1-second period. Figure C11.4-1 illustrates this for the 1-second period only using small maps of the conterminous United States that depict CR1S1UH, S1D, and S1. The reasons for using the minimum of probabilistic and deterministic spectral response accelerations, which was done previously (but not transparently) in developing the MCE ground motions maps in ASCE/SEI 7-05, are explained in the commentary of the 2003 NEHRP Recommended Provisions. In brief, deterministic ground motions provide a reasonable and practical upper-bound to design ground motions, but their use implies a somewhat higher level of collapse risk than the 1 percent probability of collapse in 50 years associated with probabilistic (risk-targeted) ground motions. In general, deterministic ground motions govern only at sites near active sources in regions of high seismicity. As defined in ASCE/SEI 7-05 Section 21.2.2, the deterministic spectral response accelerations (for Site Class B) shall not be taken as lower than 1.5g for the short periods and 0.6g for the 1-second period; hence, the ground motions on the proposed deterministic maps (Figures 22-5 and 22-6) are no lower than these values. Otherwise the ground motions on the proposed deterministic maps are 180 percent (as opposed to 150 percent in ASCE/SEI 7-05) of median spectral response accelerations, for reasons explained below in the section entitled “Deterministic Ground Motions – 84th Percentile.” Like the proposed uniform-hazard maps used in Step 1, the proposed deterministic maps represent the spectral response acceleration in the maximum direction. Step 3 – Adjust Site Class B ground motions for site condition (e.g., Site Class D) As illustrated in the bottom row of Figure C11.4-1, the third step is to multiply the spectral response accelerations resulting from Step 2 (SS and S1) by the corresponding site coefficients (Fa and Fv) from Tables 11.4-1 and 11.4-2, respectively. This step is expressed in Equation 11.4-5 for the short periods and 11.4-6 for the 1-second period, where the resulting ground motions are named risk-targeted maximum considered earthquake (MCER) spectral response accelerations and are denoted SMS and SM1, respectively. Figure C11.4-1 illustrates the step for the 1-second period only using a small map of the conterminous United States that depicts S1, an abbreviated version of Table 11.4-2, and another small map that depicts SM1. This step is the same as that in ASCE/SEI 7-05 Section 11.4.3, except that the resulting MCE spectral response accelerations (SMS and SM1) have been renamed MCER spectral response accelerations. Figures C11.4-2 and C11.4-3 are maps of the United States and California, respectively, showing values of the MCER 1- second spectral response acceleration parameter, SM1, and associated regions of Seismic Design Category, assuming Site Class D conditions. These maps illustrate MCER ground motions resulting from the three-step process described above for the 1-second period only. The design ground motions are 2/3 of these MCER ground motions as calculated using Equations 11.4-7 and 11.4-8. Summary of Design Ground Motions – 34 United States Cities Example values of the design ground motions that incorporate both USGS updates to uniform-hazard values (and hazard functions), including the new NGA relations, and the three technical changes mentioned above, are shown next. For comparison, values of design ground motions of the current standard (ASCE/SEI 7-05) and, for California sites, values of design ground motions of the 2001 California Building Code (CBC) are given. In all cases, example values are based on design ground motions, representative of Site Class D conditions (i.e., default site class). Table C11.4-1 lists the 34 city sites by region, the county (or counties) and associated populations they represent, and the latitude and longitude of the specific location of the city site. Typically, each city is the largest city of the county or metropolitan statistical area (MSA) of interest. The exception is Los Angeles County which has four city sites due to its large geographical area and associated risk. The specific location (latitude and longitude) of city sites is important for sites in high seismic regions (i.e., near an active source) since ground motions can vary greatly over relatively small distances. Example sites are selected to be coincident with the location of the hazard grid point nearest the center of the city of interest. Hazard grid points are the discrete locations at which the USGS calculates values of probabilistic and deterministic ground motions (and risk coefficients). At the time that these examples were developed, ground motions were available (from the USGS) only for these discrete locations; however, final maps and database tools such as the USGS online ground motion parameter calculator also provide values of ground motions at intermediate locations. Table C11.4-2 provides values of short-period spectral acceleration, SDS, and Table C11.4-3 provides values of 1-second spectral acceleration, SD1, for each of the 34 city sites of Table C11.4-1. Spectral acceleration values and Seismic Design Category (SDC) are given for both ASCE/SEI 7-05 provisions and changes put forth in these provisions (2009 Provisions). For California city sites, these tables also provide the corresponding values of seismic coefficients (2.5Ca, at short periods, and, Cv, at 1 second) of the 2001 California Building Code (1997 Uniform Building Code or UBC). Weighted mean values of spectral acceleration (and seismic coefficients) are calculated for each region considering the population associated with each city site in Tables 11.4-2 and 11.4-3. The following observations are made by comparing the design ground motions of these Provisions with those of ASCE/SEI 7-05 and the design coefficients of the 2001 California Building Code (CBC): 1. On a regional basis, the changes to ASCE/SEI 7-05 put forth in these Provisions result in only a slight increase or decrease in design ground motions, on average. Notable exceptions are short-period ground motions in the central and eastern United States (CEUS) for which the changes reduce design values and for certain city sites (e.g., St. Louis, Chicago, and New York) where the changes also lower the Seismic Design Category. 2. In the western region (WUS), the changes to ASCE/SEI 7-05 put forth in these Provisions result in a modest increase, or decrease, in design ground motions (plus or minus 10 percent), and generally lower seismic design values from those of 2001 CBC (1997 UBC). 3. For certain city sites (e.g., San Bernardino and San Diego), the changes to ASCE/SEI 7-05 put forth in these Provisions result in a substantial increase, or decrease, in design ground motions due primarily to changes in underlying updated USGS hazard functions. Step 1 Figure C11.4-1 Illustration of process for developing 1-second MCER Site Class D ground motions using formulas of Section 11.4.3 and associated mapped values of ground motions and risk coefficients of Chapter 22. Step 2 Figure C11.4-1 Illustration of process for developing 1-second MCER Site Class D ground motions using formulas of Section 11.4.3 and associated mapped values of ground motions and risk coefficients of Chapter 22. Step 3 Figure C11.4-1 Illustration of process for developing 1-second MCER Site Class D ground motions using formulas of Section 11.4.3 and associated mapped values of ground motions and risk coefficients of Chapter 22. Figure C11.4-1 Illustration of process for developing 1-second MCER Site Class D ground motions using formulas of Section 11.4.3 and associated mapped values of ground motions and risk coefficients of Chapter 22. S1 – Map of Site Class B ground motions Fv – Site coefficient (Table 11.4-2) SM1 – Map of MCER ground motions (Site Class D) Figure C11.4-2 Map illustrating values of the MCER 1-second spectral response acceleration parameter, SM1 (%g), and associated regions of Seismic Design Category, assuming Site Class D conditions. Figure C11.4-2 Map illustrating values of the MCER 1-second spectral response acceleration parameter, SM1 (%g), and associated regions of Seismic Design Category, assuming Site Class D conditions. Figure C11.4-3 Map illustrating values of the MCER 1-second spectral response acceleration parameter, SM1 (%g), and associated regions of Seismic Design Category, assuming Site Class D conditions, for California sites. Figure C11.4-3 Map illustrating values of the MCER 1-second spectral response acceleration parameter, SM1 (%g), and associated regions of Seismic Design Category, assuming Site Class D conditions, for California sites. Table C11.4-1 Showing Thirty-Four Cities, Site Locations (Latitude and Longitude), and Associated Counties and Populations At Risk for Which Values of Ground Motions Are Provided Table C11.4-1 Thirty-Four Cities, Site Locations (Latitude and Longitude), and Associated Counties and Populations At Risk for Which Values of Ground Motions Are Provided Table C11.4-2 Showing Comparison of Values of the Short-Period Design Ground Motion Parameter (SDS) and Corresponding Seismic Design Category (SDC) Put Forth in These Provisions with ASCE/SEI 7-05 and 1997 UBC Values for 34 City Site Locations (Assuming Default Site Class D) Table C11.4-2 Comparison of Values of the Short-Period Design Ground Motion Parameter (SDS) and Corresponding Seismic Design Category (SDC) Put Forth in These Provisions with ASCE/SEI 7-05 and 1997 UBC Values for 34 City Site Locations (Assuming Default Site Class D) Table C11.4-3 Showing Comparison of Values of the 1-Second Period Design Ground Motion Parameter (SD1) and Corresponding Seismic Design Category (SDC) Put Forth in These Provisions with ASCE/SEI 7-05 and 1997 UBC Values for 34 City Site Locations (Assuming Default Site Class D) Table C11.4-3 Comparison of Values of the 1-Second Period Design Ground Motion Parameter (SD1) and Corresponding Seismic Design Category (SDC) Put Forth in These Provisions with ASCE/SEI 7-05 and 1997 UBC Values for 34 City Site Locations (Assuming Default Site Class D) COMMENTARY TO SECTION 11.8.3 C11.8.3 Additional Geotechnical Investigation Report Requirements for Seismic Design Categories D through F. The dynamic lateral earth pressure on basement and retaining walls during the period of earthquake ground shaking is considered to be an earthquake load, E, for use in design load combinations. This dynamic earth pressure is superimposed on the preexisting static lateral earth pressure during ground shaking. The pre-existing static lateral earth pressure is considered to be an H load. While the dynamic seismic lateral earth pressures (Item 1) may be determined for design earthquake ground motions, taken as 2/3 of maximum considered earthquake ground motions, the potential for liquefaction and soil strength loss and related consequences (Items 2 and 3) must be evaluated for maximum considered earthquake ground motions because they can be catastrophic to a structure. Page intentionally left blank. Modifications to Chapter 12, Seismic Design Requirements for Building Structures TABLE 12.2-1, DESIGN COEFFICIENTS AND FACTORS FOR SEISMIC-FORCE-RESISTING SYSTEMS Revise as indicated (substantive changes are shaded, deletions are shown in strikeout, and additions are underlined): Table 12.2-1 Design Coefficients and Factors for Seismic-Force-Resisting Systems Seismic-Force-Resisting System ASCE/SEI 7-05 Section Where Detailing Requirements Are Specified Response Modification Coefficient, Ra System Overstren gth Factor, O0 g Deflection Amplificat ion Factor, Cd b Structural System Limitations and Building Height (ft) Limitc Seismic Design Category B C Dd Ed Fe A. BEARING WALL SYSTEMS 1. Special reinforced concrete shear walls 14.2 and 14.2.3.6 5 2½ 5 NL NL 160 160 100 2. Ordinary reinforced concrete shear walls 14.2 and 14.2.3.4 4 2½ 4 NL NL NP NP NP 3. Detailed plain concrete shear walls 14.2 and 14.2.3.2 2 2½ 2 NL NP NP NP NP 4. Ordinary plain concrete shear walls 14.2 and 14.2.3.1 1½ 2½ 1½ NL NP NP NP NP 5. Intermediate precast shear walls 14.2 and 14.2.3.5 4 2½ 4 NL NL 40k 40k 40k 6. Ordinary precast shear walls 14.2 and 14.2.3.3 3 2½ 3 NL NP NP NP NP 7. Special reinforced masonry shear walls 14.4 and 14.4.3 5 2½ 3½ NL NL 160 160 100 8. Intermediate reinforced masonry shear walls 14.4 and 14.4.3 3½ 2½ 2¼ NL NL NP NP NP 9. Ordinary reinforced masonry shear walls 14.4 2 2½ 1¾ NL 160 NP NP NP 10. Detailed plain masonry shear walls 14.4 2 2½ 1¾ NL NP NP NP NP 11. Ordinary plain masonry shear walls 14.4 1½ 2½ 1¼ NL NP NP NP NP 12. Prestressed masonry shear walls 14.4 1½ 2½ 1¾ NL NP NP NP NP 13. Light-framed walls sheathed with wood structural panels rated for shear resistance or steel sheets 14.1, 14.1.4.2, and 14.5 6½ 3 4 NL NL 65 65 65 14. Light-framed walls with shear panels of all other materials 14.1, 14.1.4.2, and 14.5 2 2½ 2 NL NL 35 NP NP 15. Light-framed wall systems using flat strap bracing 14.1, 14.1.4.2, and 14.5 4 2 3½ NL NL 65 65 65 16. Ordinary reinforced AAC masonry shear walls 14.4.5.4 2 2 ½ 2 NL 35 NP NP NP 17. Plain AAC masonry shear walls 14.4.5.3 1 ½ 2 ½ 1 ½ NL NP NP NP NP B. BUILDING FRAME SYSTEMS 1. Steel eccentrically braced frames, moment resisting connections at columns away from links 14.1 8 2 4 NL NL 160 160 100 2. Steel eccentrically braced frames, non-momentresisting, connections at columns away from links 14.1 7 2 4 NL NL 160 160 100 23. Special steel concentrically braced frames 14.1 6 2 5 NL NL 160 160 100 34. Ordinary steel concentrically braced frames 14.1 3¼ 2 3¼ NL NL 35j 35j NPj 45. Special reinforced concrete shear walls 14.2 and 14.2.3.6 6 2½ 5 NL NL 160 160 100 56. Ordinary reinforced concrete shear walls 14.2 and 14.2.3.4 5 2½ 4½ NL NL NP NP NP 67. Detailed plain concrete shear walls 14.2 and 14.2.3.2 2 2½ 2 NL NP NP NP NP 78. Ordinary plain concrete shear walls 14.2 and 14.2.3.1 1½ 2½ 1½ NL NP NP NP NP 89. Intermediate precast shear walls 14.2 and 14.2.3.5 5 2½ 4½ NL NL 40k 40k 40k 910. Ordinary precast shear walls 14.2 and 14.2.3.3 4 2½ 4 NL NP NP NP NP 101. Composite steel and concrete eccentrically braced frames 14.3 8 2 4 NL NL 160 160 100 112. Composite steel and concrete concentrically braced frames 14.3 5 2 4½ NL NL 160 160 100 123. Ordinary composite steel and concrete braced frames 14.3 3 2 3 NL NL NP NP NP 134. Composite steel plate shear walls 14.3 6½ 2½ 5½ NL NL 160 160 100 145. Special composite reinforced concrete shear walls with steel elements 14.3 6 2½ 5 NL NL 160 160 100 156. Ordinary composite reinforced concrete shear walls with steel elements 14.3 5 2½ 4½ NL NL NP NP NP 167. Special reinforced masonry shear walls 14.4 5½ 2½ 4 NL NL 160 160 100 178. Intermediate reinforced masonry shear walls 14.4 4 2½ 4 NL NL NP NP NP 189. Ordinary reinforced masonry shear walls 14.4 2 2½ 2 NL 160 NP NP NP 1920. Detailed plain masonry shear walls 14.4 2 2½ 2 NL NP NP NP NP 201. Ordinary plain masonry shear walls 14.4 1½ 2½ 1¼ NL NP NP NP NP 212. Prestressed masonry shear walls 14.4 1½ 2½ 1¾ NL NP NP NP NP 223. Light-framed walls sheathed with wood structural panels rated for shear resistance or steel sheets 14.1, 14.1.4.2, and 14.5 7 2½ 4½ NL NL 65 65 65 234. Light-framed walls with shear panels of all 14.1, 14.1.4.2, and 14.5 2½ 2½ 2½ NL NL 35 NP NP other materials 25. Buckling-restrained braced frames, nonmoment- resisting beamcolumn connections 14.1 7 2 5½ NL NL 160 160 100 246. Buckling-restrained braced frames, momentresisting beam-column connections 14.1 8 2½ 5 NL NL 160 160 100 257. Special steel plate shear wall 14.1 7 2 6 NL NL 160 160 100 C. MOMENT-RESISTING FRAME SYSTEMS 1. Special steel moment frames 14.1 and 12.2.5.5 8 3 5½ NL NL NL NL NL 2. Special steel truss moment frames 14.1 7 3 5½ NL NL 160 100 NP 3. Intermediate steel moment frames 12.2.5.6, 12.2.5.7, 12.2.5.8, 12.2.5.9, and 14.1 4.5 3 4 NL NL 35h,i NPh NPi 4. Ordinary steel moment frames 12.2.5.6, 12.2.5.7, 12.2.5.8, and 14.1 3.5 3 3 NL NL NPh NPh NPi 5. Special reinforced concrete moment frames 12.2.5.5 and 14.2 8 3 5½ NL NL NL NL NL 6. Intermediate reinforced concrete moment frames 14.2 5 3 4½ NL NL NP NP NP 7. Ordinary reinforced concrete moment frames 14.2 3 3 2½ NL NP NP NP NP 8. Special composite steel and concrete moment frames 12.2.5.5 and 14.3 8 3 5½ NL NL NL NL NL 9. Intermediate composite moment frames 14.3 5 3 4½ NL NL NP NP NP 10. Composite partially restrained moment frames 14.3 6 3 5½ 160 160 100 NP NP 11. Ordinary composite moment frames 14.3 3 3 2½ NL NP NP NP NP 12. Cold-formed steel – special bolted framem 14.1 3½ 3 l 3½ 35 35 35 35 35 D. DUAL SYSTEMS WITH SPECIAL MOMENT FRAMES CAPABLE OF RESISTING AT LEAST 25% OF PRESCRIBED SEISMIC FORCES 12.2.5.1 1. Steel eccentrically braced frames 14.1 8 2½ 4 NL NL NL NL NL 2. Special steel concentrically braced frames 14.1 7 2½ 5½ NL NL NL NL NL 3. Special reinforced concrete shear walls 14.2 7 2½ 5½ NL NL NL NL NL 4. Ordinary reinforced concrete shear walls 14.2 6 2½ 5 NL NL NP NP NP 5. Composite steel and concrete eccentrically braced frames 14.3 8 2½ 4 NL NL NL NL NL 6. Composite steel and concrete concentrically braced frames 14.3 6 2½ 5 NL NL NL NL NL 7. Composite steel plate 14.3 7½ 2½ 6 NL NL NL NL NL shear walls 8. Special composite reinforced concrete shear walls with steel elements 14.3 7 2½ 6 NL NL NL NL NL 9. Ordinary composite reinforced concrete shear walls with steel elements 14.3 6 2½ 5 NL NL NP NP NP 10. Special reinforced masonry shear walls 14.4 5½ 3 5 NL NL NL NL NL 11. Intermediate reinforced masonry shear walls 14.4 4 3 3½ NL NL NP NP NP 12. Buckling-restrained braced frame 14.1 8 2½ 5 NL NL NL NL NL 13. Special steel plate shear walls 14.1 8 2½ 6½ NL NL NL NL NL E. DUAL SYSTEMS WITH INTERMEDIATE MOMENT FRAMES CAPABLE OF RESISTING AT LEAST 25% OF PRESCRIBED SEISMIC FORCES 12.2.5.1 1. Special steel concentrically braced framesf 14.1 6 2½ 5 NL NL 35 NP NPh,k 2. Special reinforced concrete shear walls 14.2 6½ 2½ 5 NL NL 160 100 100 3. Ordinary reinforced masonry shear walls 14.4 3 3 2½ NL 160 NP NP NP 4. Intermediate reinforced masonry shear walls 14.4 3½ 3 3 NL NL NP NP NP 5. Composite steel and concrete concentrically braced frames 14.3 5½ 2½ 4½ NL NL 160 100 NP 6. Ordinary composite braced frames 14.3 3½ 2½ 3 NL NL NP NP NP 7. Ordinary composite reinforced concrete shear walls with steel elements 14.3 5 3 4½ NL NL NP NP NP 8. Ordinary reinforced concrete shear walls 14.2 5½ 2½ 4½ NL NL NP NP NP F. SHEAR WALLFRAME INTERACTIVE SYSTEM WITH ORDINARY REINFORCED CONCRETE MOMENT FRAMES AND ORDINARY REINFORCED CONCRETE SHEAR WALLS 12.2.5.10 and 14.2 4½ 2½ 4 NL NP NP NP NP G. CANTILEVERED COLUMN SYSTEMS DETAILED TO CONFORM TO THE REQUIREMENTS FOR: 12.2.5.2 1. Special steel moment frames 12.2.5.5 and 14.1 2½ 1¼ 2½ 35 35 35 35 35 2. Intermediate steel moment frames 14.1 1½ 1¼ 1½ 35 35 35h NPh,i NPh,i 3. Ordinary steel moment frames 14.1 1¼ 1¼ 1¼ 35 35 NP NPh,i NPh,i 4. Special reinforced concrete moment frames 12.2.5.5 and 14.2 2½ 1¼ 2½ 35 35 35 35 35 5. Intermediate concrete moment frames 14.2 1½ 1¼ 1½ 35 35 NP NP NP 6. Ordinary concrete moment frames 14.2 1 1¼ 1 35 NP NP NP NP 7. Timber frames 14.5 1½ 1½ 1½ 35 35 35 NP NP H. STEEL SYSTEMS NOT SPECIFICALLY DETAILED FOR SEISMIC RESISTANCE, EXCLUDING CANTILEVER COLUMN SYSTEMS 14.1 3 3 3 NL NL NP NP NP aResponse modification coefficient, R, for use throughout the standard. Note R reduces forces to a strength level, not an allowable stress level. bReflection amplification factor, Cd, for use in Sections 12.8.6, 12.8.7, and 12.9.2 cNL = Not Limited and NP = Not Permitted. For metric units use 30.5 m for 100 ft and use 48.8 m for 160 ft. Heights are measured from the base of the structure as defined in Section 11.2. dSee Section 12.2.5.4 for a description of building systems limited to buildings with a height of 240 ft (73.2 m) or less. eSee Section 12.2.5.4 for building systems limited to buildings with a height of 160 ft (48.8 m) or less. fOrdinary moment frame is permitted to be used in lieu of intermediate moment frame for Seismic Design Categories B or C. gThe tabulated value of the overstrength factor, O0, is permitted to be reduced by subtracting one-half for structures with flexible diaphragms, but shall not be taken as less than 2.0 for any structure. hSee Sections 12.2.5.6 and 12.2.5.7 for limitations for steel OMFs and IMFs in structures assigned to Seismic Design Category D or E. iSee Sections 12.2.5.8 and 12.2.5.9 for limitations for steel OMFs and IMFs in structures assigned to Seismic Design Category F. jSteel ordinary concentrically braced frames are permitted in single-story buildings up to a height of 60 ft (18.3 m) where the dead load of the roof does not exceed 20 psf (0.96 kN/m2) and in penthouse structures. l Alternatively, the seismic load effect with overstrength, Em, can be based on the expected strength determined in accordance with AISI S110. m Cold-formed steel – special bolted moment frames shall be limited to one story in height in accordance with AISI S110. TABLE 12.6-1, PERMITTED ANALYTICAL PROCEDURES Replace with the following: Table 12.6-1 Permitted Analytical Procedures Seismic Design Category Structural Characteristics Equivalent Lateral Force Analysis Section 12.8 Modal Response Spectrum Analysis Section 12.9 Seismic Response History Procedures Chapter 16 B, C All structures P P P D, E, F Regular structures not exceeding 160 feet in height and all structures of light frame construction P P P Regular structures equal to or exceeding 160 feet in height with T < 3.5 Ts P P P Irregular structures not exceeding 160 feet in height and having only horizontal irregularities type 2, 3, 4, or 5 of Table 12.3- 1 or vertical irregularities type 4, 5a or 5b of Table 12.3-2 P P P All other structures NP P P Note: P – Permitted; NP – Not permitted. SECTION 12.8.7, P-DELTA LIMIT Replace with the following: 12.8.7 P-delta Limit. Stability coefficient, ., as determined for each level of the structure by the following equation, shall not exceed 0.10: (12.8-16) where: Px = the total vertical design load at and above Level x. Where calculating the vertical design load for purposes of determining P-delta effects, the individual load factors need not exceed 1.0. . = the design story drift calculated in accordance with Section 12.8.6. I = the occupancy importance factor determined in accordance with Section 11.5.1. Vx = the seismic shear force acting between Level x and x - 1. hsx = the story height below Level x. Cd = the deflection amplification factor from Table 12.2-1 EXCEPTION: The stability coefficient, ., shall be permitted to exceed 0.10 if either of the following applies: Equation Equation 1. The resistance to lateral forces is determined to increase continuously in a monotonic nonlinear static (pushover) analysis according to ASCE/SEI 41 Section 3.3.3.3.2 using Sa defined as a MCER spectral response acceleration according to the Provisions at the effective fundamental period. Modeling and analysis shall conform to ASCE/SEI 41 Section 3.3.3, except that the analysis shall be done for seismic actions occurring simultaneously with the effects of dead load in combination with not less than 25 percent of the required design live loads, reduced as permitted for the area of a single floor. Degradation shall be modeled and P-delta effects shall be included in the analysis. A review of the nonlinear static analysis shall be performed by an independent team having experience in seismic analysis methods and the theory and application of nonlinear seismic analysis and structural behavior under earthquake loading. The review team shall be composed of at least two members including at least one registered design professional. 2. Compliance with the provisions of the nonlinear response history procedure in Chapter 16 is demonstrated. SECTION 12.11.2.2.1, TRANSFER OF ANCHORAGE FORCES INTO DIAPHRAGM Replace with the following: 12.11.2.2.1 Transfer of Anchorage Forces into Diaphragm. Diaphragms shall be provided with continuous ties or struts between diaphragm chords to distribute these anchorage forces into the diaphragm. EXCEPTION: In buildings with diaphragms of light-frame construction, continuous cross-ties are not required provided all of the following are satisfied: 1. The unsupported height of the wall does not exceed 12 feet, 2. Anchorages are spaced no more than 4 feet on center, 3. The length of the diaphragm in the direction parallel to the wall being anchored does not exceed 2.5 times the length of the diaphragm in the orthogonal direction, and 4. The anchorage connection extends far enough into the diaphragm to transfer the anchorage force into the diaphragm. Diaphragm connections shall be positive, mechanical, or welded. Added chords are permitted to be used to form subdiaphragms to transmit the anchorage forces to the main continuous cross-ties. The maximum length-to-width ratio of the structural subdiaphragm shall be 2.5 to 1. Connections and anchorages capable of resisting the prescribed forces shall be provided between the diaphragm and the attached components. Connections shall extend into the diaphragm a sufficient distance to develop the force transferred into the diaphragm. SECTION 12.11.2.2.3, WOOD DIAPHRAGMS Replace with the following: 12.11.2.2.3 Wood Diaphragms. In wood diaphragms, the continuous ties shall be in addition to the diaphragm sheathing. EXCEPTION: Where continuous cross-ties are not required by Section 12.11.2.2.1 and the anchorage connections extend into the diaphragm a sufficient distance to develop the force transferred into the diaphragm sheathing. Anchorage shall not be accomplished by use of toenails or nails subject to withdrawal nor shall wood ledgers or framing be used in cross-gain bending or cross-grain tension. The diaphragm sheathing shall not be considered effective as providing the ties or struts required by this section. SECTION 12.14.7.5.1, TRANSFER OF ANCHORAGE FORCES INTO DIAPHRAGM Replace with the following: 12.14.7.5.1 Transfer of Anchorage Forces into Diaphragm. Diaphragms shall be provided with continuous ties or struts between diaphragm chords to distribute these anchorage forces into the diaphragm. EXCEPTION: In buildings with diaphragms of light-framed construction, continuous cross-ties are not required provided all of the following are satisfied: 1. The unsupported height of the wall does not exceed 12 feet, 2. Anchorages are spaced no more than 4 feet on center, 3. The length of the diaphragm in the direction parallel to the wall being anchored does not exceed 2.5 times the length of the diaphragm in the orthogonal direction, and 4. The connection extends far enough into the diaphragm to transfer the anchorage force into the diaphragm. Added chords are permitted to be used to form subdiaphragms to transmit the anchorage forces to the main continuous cross-ties. The maximum length-to-width ratio of the structural subdiaphragm shall be 2.5 to 1. Connections and anchorages capable of resisting the prescribed forces shall be provided between the diaphragm and the attached components. Connections shall extend into the diaphragm a sufficient distance to develop the force transferred into the diaphragm. SECTION 12.14.7.5.2, WOOD DIAPHRAGMS Replace with the following: 12.14.7.5.2 Wood Diaphragms. In wood diaphragms, the continuous ties shall be in addition to the diaphragm sheathing. EXCEPTION: Where continuous cross-ties are not required by Section 12.14.7.5.1 and the anchorage connections extend into the diaphragm a sufficient distance to develop the force transferred into the diaphragm sheathing. Anchorage shall not be accomplished by use of toenails or nails subject to withdrawal nor shall wood ledgers or framing be used in cross-gain bending or cross-grain tension. The diaphragm sheathing shall not be considered effective as providing the ties or struts required by this section. SECTION 12.14.8.1, SEISMIC BASE SHEAR Revise, in part, to read as follows: . . . In calculating SDS, Ss shall be in accordance with Section 11.4.3, but need not be taken larger than 1.5. [Remainder of section stays the same.] Commentary to Chapter 12 Modifications COMMENTARY TO SECTION 12.6 C12.6 ANALYSIS SELECTION PROCEDURE Table 12.6-1 applies only to buildings without seismic isolation (Chapter 17) or passive energy devices (Chapter 18). The four basic procedures addressed in Table 12.6-1 are equivalent lateral force (ELF) analysis (Section 12.8), modal response spectrum (MRS) analysis (Section 12.9), linear response history (LRH) analysis, and nonlinear response history (NRH) analysis. Requirements for performing response history analysis are provided in Chapter 16. Nonlinear static (pushover) analysis is not provided as an “approved” analysis procedure in ASCE/SEI 7-05. The value of Ts = SD1/SDS depends on the site class because SDS and SD1 include such effects. When ELF is not allowed, the analysis must be performed using modal response spectrum or response history analysis. ELF is not allowed for buildings with the listed irregularities because it assumes a gradually varying distribution of mass and stiffness along the height and negligible torsional response. The 3.5Ts limit recognizes that higher modes are more significant in taller buildings (Lopez and Cruz, 1996; Chopra, 2007), such that the ELF method may underestimate the design base shear and may not predict correctly the vertical distribution of seismic forces. Table C12.6-1 demonstrates that 3.5Ts generally increases as ground motion intensity increases and as soils become softer. Assuming that the fundamental building period is about 0.1 times the number of stories, the maximum building height for which the ELF applies ranges from about 10 stories for low seismic hazard sites with firm soil to 30 stories for high seismic hazard sites with soft soil. Since this trend was not intended, the modification to Section 12.6 adds a height limit of 160 feet. Table C12.6-1 Values of 3.5TS for Various Cities and Various Site Classes Location Ss (g) S1 (g) 3.5Ts (seconds) for Site Class A&B C D E Denver 0.219 0.057 0.91 1.29 1.37 1.07 Boston 0.275 0.067 0.85 1.21 1.30 1.03 New York City 0.359 0.070 0.68 0.97 1.08 0.93 Las Vegas 0.582 0.179 1.08 1.50 1.68 1.89 St. Louis 0.590 0.169 1.00 1.40 1.60 1.81 San Diego 1.128 0.479 1.31 1.73 1.99 2.91 Memphis 1.341 0.368 0.96 1.38 1.59 2.25 Charleston 1.414 0.348 0.86 1.25 1.47 2.08 Seattle 1.448 0.489 1.18 1.55 1.78 2.63 San Jose 1.500 0.600 1.40 1.82 2.10 2.12 Salt Lake City 1.672 0.665 1.39 1.81 2.09 3.10 COMMENTARY TO SECTION 12.8.7 C12.8.7 P-delta Limit. ASCE/SEI 7-05 allows amplified forces to be used in a linear elastic analysis where the stability coefficient, ., exceeds 0.10. By comparison, FEMA 350 requires explicit modeling of P-delta effects for steel momentresisting frames where . exceeds approximately 0.04. Where the tangent stiffness of the structure may become negative, dynamic displacement demands can increase significantly (Gupta and Krawinkler, 2000). Structures with . not greater than 0.10 generally are expected to have a positive tangent stiffness, depending on the progression of plastic hinging and strain hardening. The 2009 Provisions allows structures to exceed this limit only if a nonlinear static analysis including P-delta effects demonstrates that the tangent stiffness remains positive up to the target displacement computed for the MCER or if nonlinear dynamic analysis demonstrates adequate resistance to instability. The occupancy importance factor, I, is inserted into Equation 12.8-16 to correct an error in ASCE/SEI 7. In this way, the stability coefficient is based on the elastic stiffness of the system. ADDITIONAL REFERENCES FOR CHAPTER 12 COMMENTARY Federal Emergency Management Agency. 2000. Recommended Seismic Design Criteria for New Steel Moment-Frame Buildings, FEMA 350. Prepared for FEMA by the SAC Joint Venture. Federal Emergency Management Agency, Washington, D.C. Gupta, A., and H. Krawinkler. 2000. “Dynamic P-delta effects for flexible inelastic steel structures,” ASCE Journal of Structural Engineering, 126(1):145-154. Modifications to Chapter 13, Seismic Design Requirements for Nonstructural Elements SECTION 13.6.5.5, ADDITIONAL REQUIREMENTS [FOR COMPONENT SUPPORTS] Replace Item 6f with the following: Attachments into concrete utilize anchors that have not been prequalified for seismic applications in accordance with ACI 355.2. SECTION 13.6.8.2, FIRE PROTECTION SPRINKLER SYSTEMS IN SEISMIC DESIGN CATEGORY C Replace with the following: 13.6.8.2 Fire Protection Sprinkler Systems. Fire protection sprinkler systems designed and constructed in accordance with NFPA 13 shall be deemed to meet the other requirements of this section. SECTION 13.6.8.3, FIRE PROTECTION SPRINKLER SYSTEMS IN SEISMIC DESIGN CATEGORIES D THROUGH F Delete this section and renumber remaining sections. Commentary to Chapter 13 Modifications COMMENTARY TO SECTION 13.6.5.5 C13.6.5.5 Additional Requirements [for Component Supports]. As reflected in this section of the standard and in the footnote to Table 13.6-1, vibration isolated equipment with snubbers is subject to amplified loads as a result of dynamic impact. Use of expansion anchors for non-vibration isolated mechanical equipment rated over 10 hp is prohibited based on experience with older anchor types. The ASCE 7 Seismic Subcommittee is considering a proposal that also would exempt anchors qualified by simulated seismic testing and long-term vibration testing. The previous language in Item 6f was intended to identify anchor types that would be considered non-ductile. The previous requirement has been superseded by requirements for qualification that include checks for ductility and good performance in earthquake conditions. COMMENTARY TO SECTION 13.6.8.2 C13.6.8.2 Fire Protection Sprinkler Systems. NFPA 13-2007 applies to Seismic Design Categories C, D, E, and F. The lateral design procedures of NFPA 13-2007 have been revised for consistency with the ASCE/SEI 7-05 design approach while retaining traditional sprinkler system design concepts. Using conservative upper-bound values of the various design parameters, a single lateral force coefficient, Cp, was developed. It is a function of the mapped short period response parameter Ss. Stresses in the pipe and connections are controlled by limiting the maximum reaction at bracing points as a function of pipe diameter. In Seismic Design Category C, the prescriptive requirements of NFPA 13-2007, using a default lateral force of 50 percent of the weight of the water-filled pipe, provide a conservative design, although application of the NFPA sway bracing calculation may produce a lower design lateral force. Page intentionally left blank. Modifications to Chapter 14, Material Specific Seismic Design and Detailing Requirements SECTION 14.1, STEEL Replace with the following: 14.1 STEEL Structures, including foundations, constructed of steel to resist seismic loads shall be designed and detailed in accordance with this standard including the reference documents and additional requirements provided in this section. 14.1.1 Reference Documents. The design, construction, and quality of steel components that resist seismic forces shall conform to the applicable requirements of the following as amended herein: 1. AISC 360 2. AISC 341 3. AISI NAS 4. AISI S110 5. AISI-GP 6. AISI-PM 7. AISI Lateral 8. AISI WSD 9. ASCE 19 10. ASCE 8 11. SJI Tables 14.1.1.1 Modifications to AISC 341-05. The text of AISC 341 shall be modified as indicated in Sections 14.1.1.1.1 and 14.1.1.1.2. Italics are used for text to indicate requirements that differ from AISC 341. 14.1.1.1.1 Replace Section 15.7 with the following: 15.7 Beam-to-Column Connections Where a brace or gusset plate connects to both members at a beam-to-column connection, the connection shall conform to one of the following: (1) The connection shall accommodate the required rotation at a minimum story drift of 2.5 percent of the story height or (2) The connection shall be designed to resist a moment equal to the lesser of the following: (i) A moment corresponding to 1.1RyFyZ (LRFD) or (1.1/1.5)RyFyZ (ASD), as appropriate, of the beam. (ii) A moment corresponding to S1.1RyFyZ (LRFD) or S[(1.1/1.5)RyFyZ] (ASD), as appropriate, of the column. This moment shall be considered in combination with the required strength of the brace connection and beam connection, including amplified diaphragm collector forces. 1.4.1.1.1.2 Add new Section 16.7 as follows: 16.7 Beam-to-Column Connections Where a brace or gusset plate connects to both members at a beam-to-column connection, the connection shall conform to one of the following: (1) The connection shall accommodate the required rotation at a minimum story drift of 2.5 percent of the story height or (2) The connection shall be designed to resist a moment equal to the lesser of the following: (i) A moment corresponding to 1.1RyFyZ (LRFD) or (1.1/1.5)RyFyZ (ASD), as appropriate, of the beam. (ii) A moment corresponding to S1.1RyFyZ (LRFD) or S[(1.1/1.5)RyFyZ] (ASD), as appropriate, of the column. This moment shall be considered in combination with the required strength of the brace connection and beam connection, including amplified diaphragm collector forces. 14.1.2 Seismic Design Categories B and C. Steel structures assigned to Seismic Design Category B or C shall be of any construction permitted by the reference documents in Section 14.1.1. An R factor as set forth in Table 12.2-1 is permitted where the structure is designed and detailed in accordance with the requirements of AISC 341 for structural steel buildings, AISI S110 for cold-formed steel construction, or AISI Lateral for light-framed cold-formed steel construction. Systems not detailed in accordance with AISC 341, AISI S110, or AISI Lateral shall use the R factor designated for “Structural steel systems not specifically detailed for seismic resistance” in Table 12.2-1. 14.1.3 Seismic Design Categories D through F. Steel structures assigned to Seismic Design Category D, E, or F shall be designed and detailed in accordance with AISC 341 for structural steel, AISI S110 for cold-formed steel construction, or AISI Lateral for light-framed cold-formed steel construction. 14.1.4 Cold-formed Steel. The design of cold-formed carbon or low-alloy steel to resist seismic loads shall be in accordance with the requirements of AISI NAS, AISI S110 and the design of cold-formed stainless steel structural members to resist seismic loads shall be in accordance with the requirements of ASCE 8. 14.1.4.1 Modifications to AISI S110 (2007 edition). The text of AISI S110 shall be modified as indicated in Sections 14.1.4.1.1 through 14.1.2.1.5. Italics are used for text within Sections 14.1.4.1.1 through 14.1.2.1.5 to indicate requirements that differ from AISI S110. 14.1.4.1.1 AISI S110, Section D1. Revise Section D1 to read as follows: D1 Cold-Formed Steel Special Bolted Moment Frames (CFS-SBMF) Cold-formed steel–special bolted moment frames (CFS-SBMF) systems shall withstand inelastic deformations through friction and bearing at their bolted connections. Beams, columns, and connections shall satisfy the requirements in this section. CFS-SBMF systems shall be limited to one-story structures, no greater than 35 feet in height, without column splices and satisfying the requirements in this section. The SBMF shall engage all columns supporting the roof or floor above. The single size beam and single size column with the same bolted moment connection detail shall be used for each frame. The frame is to be supported on a level floor or foundation. 14.1.4.1.2 AISI S110, Section D1.1.1. Revise Section D1.1.1 to read as follows: D1.1.1 Connection Limitations Beam-to-column connections in CFS-SBMF systems shall be bolted connections with snug-tight highstrength bolts. The bolt spacing and edge distance shall be in accordance with the limits of AISI S100, Section E3. The 8-bolt configuration shown in Table D1-1 shall be used. The faying surfaces of the beam and column in the bolted moment connection region shall be free of any lubricants or debris. 14.1.4.1.3 AISI S110, Section D1.2.1. Revise Section D1.2.1 to read as follows: D1.2.1 Beam Limitations In addition to the requirements of Section D1.2.3, beams in CFS-SBMF systems shall be ASTM A653 Gr. 55 galvanized steel cold-formed C-sections members with lips, and designed in accordance with Chapter C of AISI S100. The beam depth shall be between 12 in (305 mm) and 20 in (508 mm). The flat depth-to-thickness ratio of the web shall not exceed 6.18 . 14.1.4.1.4 AISI S110, Section D1.2.2. Revise Section D1.2.2 to read as follows: D1.2.2 Column Limitations In addition to the requirements of D1.2.3, columns in CFS-SBMF systems shall be ASTM A500 Gr. B coldformed hollow structural section (HSS) members painted with a standard industrial finished surface, and designed in accordance with Chapter C of AISI S100. The column depth shall be between 8 in (203 mm) and 12 in (305 mm). The flat depth-to-thickness ratio shall not exceed 1.40 . Equation Equation Equation E / Fy 14.1.4.1.5 AISI S110, Section D1.3. Revise Section D1.3 to read as follows: D1.3 Design Story Drift Where the applicable building code does not contain design coefficients for CSF-SBMF systems, the provisions of Appendix 1 shall apply. The design story drift shall not exceed 0.03h, unless approved by authority having jurisdiction. In no case shall the design story drift exceed 0.05h. For structures having a period less than TS, as defined in the applicable building code, alternate methods of computing . shall be permitted, provided such alternate methods are acceptable to the authority having jurisdiction. [Remainder of Section 14.1 is unchanged.] SECTION 14.2.2, MODIFICATIONS TO ACI 318 Replace with the following: 14.2.2 Modifications to ACI 318. The text of ACI 318 shall be modified as indicated in Sections 14.2.2.1 through 14.2.2.9. Italics are used for text within Sections 14.2.2.1 through 14.2.2.9 to indicate provisions that differ from ACI 318. 14.2.2.1 Definitions. Add the following definitions to Section 2.2. DETAILED PLAIN CONCRETE STRUCTURAL WALL: A wall complying with the requirements of Chapter 22. ORDINARY PRECAST STRUCTURAL WALL: A precast wall complying with the requirements of Chapters 1 through 18. WALL PIER: A wall segment with a horizontal length-to-thickness ratio of at least 2.5, but not exceeding 6, whose clear height is at least two times its horizontal length. 14.2.2.2 ACI 318, Section 7.10. Modify Section 7.10 by revising Section 7.10.5.6 to read as follows: 7.10.5.6 Where anchor bolts are placed in the top of columns or pedestals, the bolts shall be enclosed by lateral reinforcement that also surrounds at least four vertical bars of the column or pedestal. The lateral reinforcement shall be distributed within 5 in. of the top of the column or pedestal, and shall consist of at least two No.4 or three No.3 bars. In structures assigned to Seismic Design Categories C, D, E or F, the ties shall have a hook on each free end that complies with 7.1.4. 14.2.2.3 Scope. Modify Section 21.1.1.3 to read as follows: 21.1.1.3 All members shall satisfy requirements of Chapters 1 to 19 and 22. Structures assigned to SDC B, C, D, E, or F also shall satisfy 21.1.1.4 through 21.1.1.8, as applicable, except as modified by the requirements of Chapters 14 and 15 of this document. 14.2.2.4 Intermediate Precast Structural Walls: Modify Section 21.4 by renumbering Section 21.4.3 to Section 21.4.4 and adding new Sections 21.4.3 and 21.4.5, to read as follows: 21.4 Connections that are designed to yield shall be capable of maintaining 80 percent of their design strength at the deformation induced by design displacement, or shall use type 2 mechanical splices. 21.4.4 Elements of the connection that are not designed to yield shall develop at least 1.5 Sy. 21.4.5 Wall piers not designed as part of a moment frame shall have transverse reinforcement designed to resist the shear forces determined from Section 21.3.3. Spacing of transverse reinforcement shall not exceed 8 in. Transverse reinforcement shall be extended beyond the pier clear height for at least 12 in. EXCEPTIONS: The preceding requirement need not apply in the following situations: 1. Wall piers that satisfy Section 21.13. 2. Wall piers along a wall line within a story where other shear wall segments provide lateral support to the wall piers and such segments have a total stiffness of at least six times the sum of the in-plane stiffnesses of all the wall piers. Wall segments with a horizontal length-to-thickness ratio less than 2.5 shall be designed as columns. 14.2.2.5 Wall Piers and Wall Segments. Modify Section 21.9 by adding a new Section 21.9.10 to read as follows: 21.9.10 Wall Piers and Wall Segments in Special Structural Walls. 21.9.10.1 Wall piers not designed as a part of a special moment-resisting frame shall have transverse reinforcement designed to satisfy the requirements in Section 21.9.10.2. EXCEPTIONS: 1. Wall piers that satisfy Section 21.13. 2. Wall piers along a wall line within a story where other shear wall segments provide lateral support to the wall piers, and such segments have a total stiffness of at least six times the sum of the in-plane stiffnesses of all the wall piers. 21.9.10.2 Transverse reinforcement with seismic hooks at both ends shall be designed to resist the shear forces determined from Section 21.6.5.1. Spacing of transverse reinforcement shall not exceed 6 in. (152 mm). Transverse reinforcement shall be extended beyond the pier clear height for at least 12 in. (304 mm). 21.9.10.3 Wall segments with a horizontal length-to-thickness ratio less than 2.5 shall be designed as columns. 14.2.2.6 Special Precast Structural Walls. Modify Section 21.10.2 to read as follows: 21.10.2 Special structural walls constructed using precast concrete shall satisfy all the requirements of Section 21.9 in addition to 21.4 as modified in Section 14.2.2.7. 14.2.2.7 Foundations. Modify Section 21.12.1.1 to read as follows: 21.12.1.1 Foundations resisting earthquake-induced forces or transferring earthquake-induced forces between structure and ground shall comply with requirements of Section 21.12 and other applicable code provisions unless modified by Sections 12.1.5, 12.13 or 14.2 of ASCE/SEI 7-05. 14.2.2.8 Detailed Plain Concrete Shear Walls. Modify Section 22.6 by adding a new Section 22.6.7 to read: 22.6.7 Detailed Plain Concrete Shear Walls. 22.6.7.1 Detailed plain concrete shear walls are walls conforming to the requirements for ordinary plain concrete shear walls and 22.6.7.2 22.6.7.2 Reinforcement shall be provided as follows: a. Vertical reinforcement of at least 0.20 in.2 (129 mm2) in cross-sectional area shall be provided continuously from support to support at each corner, at each side of each opening, and at the ends of walls. The continuous vertical bar required beside an opening is permitted to substitute for the No. 5 bar required by Section 22.6.6.5. b. Horizontal reinforcement at least 0.20 in.2 (129 mm2) in cross-sectional area shall be provided: 1. Continuously at structurally connected roof and floor levels and at the top of walls. 2. At the bottom of load-bearing walls or in the top of foundations where doweled to the wall 3. At a maximum spacing of 120 in. (3048 mm). Reinforcement at the top and bottom of openings, where used in determining the maximum spacing specified in Item 3 in the preceding text, shall be continuous in the wall. 14.2.2.9 Strength Requirements for Anchors: Modify Section D.4 by adding a new exception at the end of Section D.4.2.2 to read as follows: EXCEPTION: If Nb is determined using Equation D-7, the concrete breakout strength of Section D.4.2 shall be considered satisfied by the design procedure of Sections D.5.2 and D.6.2 without the need for testing regardless of anchor bolt diameter and tensile embedment. SECTIONS 14.2.3, ADDITIONAL DETAILING REQUIREMENTS FOR CONCRETE PILES, AND 14.2.3.1, CONCRETE PILE REQUIREMENTS FOR SDC C Replace with the following: 14.2.3 Additional Detailing Requirements for Concrete Piles. In addition to the foundation requirements set forth in ACI 318 Sections 12.1.5, 12.13 and 21.12, design, detailing and construction of concrete piles shall conform to the provisions of this section. 14.2.3.1 Concrete Pile Requirements for Seismic Design Category C. Concrete piles in structures assigned to Seismic Design Category C shall comply with the requirements of this section. 14.2.3.1.1 Anchorage of Piles. All concrete piles and concrete filled pipe piles shall be connected to the pile cap by embedding the pile reinforcement in the pile cap for a distance equal to the development length as specified in ACI 318 as modified by Section 14.2.2 of this standard or by the use of field-placed dowels anchored in the concrete pile. For deformed bars, the development length is the full development length for compression or tension, in the case of uplift, without reduction in length for excess area. Hoops, spirals, and ties shall be terminated with seismic hooks as defined in ACI 318 Section 2.2. Where a minimum length for reinforcement or the extent of closely spaced confinement reinforcement is specified at the top of the pile, provisions shall be made so that those specified lengths or extents are maintained after pile cut-off. SECTION 14.2.3.2, CONCRETE PILE REQUIREMENTS FOR SEISMIC DESIGN CATEGORIES D THROUGH F Replace Sections 14.2.3.2.1 through 14.2.3.2.5 with the following: 14.2.3.2.1 Site Class E or F Soil. Where concrete piles are used in Site Class E or F, they shall have transverse reinforcement in accordance with ACI 318 Sections 21.6.4.2 through 21.6.4.4 within seven pile diameters of the pile cap and the interfaces between strata that are hard or stiff and strata that are liquefiable or are composed of soft to medium stiff clay. 14.2.3.2.2 Nonapplicable ACI 318 Sections for Grade Beam and Piles. ACI 318 Section 21.12.3.3 need not apply where grade beams have the required strength to resist the forces from the load combinations with overstrength factor of Section 12.4.3.2 or 12.14.3.2.2. ACI 318 Section 21.12.4.4(a) need not apply to concrete piles, and Section 21.12.4.4(b) need not apply to precast, prestressed concrete piles. 14.2.3.2.3 Reinforcement for Uncased Concrete Piles (SDC D through F). Reinforcement shall be provided where required by analysis. For uncased cast-in-place drilled or augered concrete piles, a minimum of four longitudinal bars with a minimum longitudinal reinforcement ratio of 0.005 and transverse reinforcement in accordance with ACI 318 Sections 21.6.4.2 through 21.6.4.4 shall be provided throughout the minimum reinforced length of the pile as defined below starting at the top of the pile. The longitudinal reinforcement shall extend beyond the minimum reinforced length of the pile by the tension development length. The minimum reinforced length of the pile shall be taken as the greater of: 1. One-half of the pile length; 2. A distance of 10 ft (3 m); 3. Three times the pile diameter; 4. The flexural length of the pile which shall be taken as the length of from the bottom of the pile cap to a point where the concrete section cracking moment multiplied by a resistance factor 0.4 exceeds the required factored moment at that point. In addition, for piles located in Site Class E or F, longitudinal reinforcement and transverse confinement reinforcement, as described above, shall extend the full length of the pile. Where transverse reinforcement is required, transverse reinforcing ties shall be a minimum of No. 3 bars for up to 20-in.- diameter (300 mm) piles and No.4 bars for piles of larger diameter. In Site Classes A through D, longitudinal reinforcement and transverse confinement reinforcement, as defined above, shall extend a minimum of seven times the pile diameter above and below the interfaces of soft to medium stiff clay or liquefiable strata except that transverse reinforcing ties not located within the minimum reinforced length shall be permitted to use a transverse spiral reinforcement ratio of not less than one-half of that required in ACI 318 Section 21.6.4.4(a). Spacing of transverse reinforcement not located within the minimum reinforced length is permitted to be increased, but shall not exceed the least of the following: 1. 12 longitudinal bar diameters; 2. One-half the pile diameter; 3. 12 in. (305 mm). 14.2.3.2.4 Reinforcement for Metal-Cased Concrete Piles (SDC D through F). Reinforcement requirements are the same as for uncased concrete piles. EXCEPTION: Spiral-welded metal-casing of a thickness not less than No. 14 gauge can be considered as providing concrete confinement equivalent to the closed ties or equivalent spirals required in an uncased concrete pile, provided that the metal casing is adequately protected against possible deleterious action due to soil constituents, changing water levels, or other factors indicated by boring records of site conditions. 14.2.3.2.5 Reinforcement for Precast Concrete Piles (SDC D through F). Transverse confinement reinforcement consisting of closed ties or equivalent spirals shall be provided in accordance with ACI 318 Sections 21.6.4.2 through 21.6.4.4 for the full length of the pile. EXCEPTION: In other than Site Classes E or F, the specified transverse confinement reinforcement shall be provided within three pile diameters below the bottom of the pile cap, but it shall be permitted to use a transverse reinforcing ratio of not less than one-half of that required in ACI 318 Section 21.6.4.4(a) throughout the remainder of the pile length. [Remainder of Section 14.2.3.2 is unchanged.] NEW SECTION 14.2.4, ACCEPTANCE CRITERIA FOR SPECIAL PRECAST STRUCTURAL WALLS BASED ON VALIDATION TESTING Add the following new section: 14.2.4 Acceptance Criteria for Special Precast Structural Walls Based on Validation Testing 14.2.4.1 Notation Symbols additional to those in ACI 318 Chapter 2 are defined. Emax = maximum lateral resistance of test module determined from test results (forces or moments) En = nominal lateral resistance of test module calculated using specified geometric properties of test members, specified yield strength of reinforcement, specified compressive strength of concrete, a strain compatibility analysis or deformation compatibility analysis for flexural strength and a strength reduction factor f of 1.0 Ent = calculated lateral resistance of test module using the actual geometric properties of test members, the actual strengths of reinforcement, concrete, and coupling devices, obtained by testing per Sections 14.2.4.7.7, 14.2.4.7.8, and 14.2.4.7.9, and a strength reduction factor f of 1.0 . = drift ratio ß = relative energy dissipation ratio 14.2.4.2 Definitions Definitions additional to those in ACI 318 Chapter 2 are defined. 14.2.4.2.1 Coupling Elements. Devices or beams connecting adjacent vertical boundaries of structural walls and used to provide stiffness and energy dissipation for the connected assembly greater than the sum of those provided by the connected walls acting as separate units. 14.2.4.2.2 Drift Ratio. Total lateral deformation of the test module divided by the height of the test module. 14.2.4.2.3 Global Toughness. The ability of the entire lateral force resisting system of the prototype structure to maintain structural integrity and continue to carry the required gravity load at the maximum lateral displacements anticipated for the ground motions of the maximum considered earthquake. 14.2.4.2.4 Prototype Structure. The concrete wall structure for which acceptance is sought. 14.2.4.2.5 Relative Energy Dissipation Ratio. Ratio of actual to ideal energy dissipated by test module during reversed cyclic response between given drift ratio limits, expressed as the ratio of the area of the hysteresis loop for that cycle to the area of the circumscribing parallelograms defined by the initial stiffnesses during the first cycle and the peak resistances during the cycle for which the relative energy dissipation ratio is calculated. See Section 14.2.4.9.1.3. 14.2.4.2.5 Test Module. Laboratory specimen representing the critical walls of the prototype structure. See Section 14.2.4.5. 14.2.4.3 Scope and General Requirements 14.2.4.3.1 These provisions define minimum acceptance criteria for new precast structural walls, including coupled precast structural walls, designed for regions of high seismic risk or for structures assigned to high seismic performance or design categories, where acceptance is based on experimental evidence and mathematical analysis. 14.2.4.3.2 These provisions are applicable to precast structural walls, coupled or uncoupled, with height to length, hw/lw, ratios equal to or greater than 0.5. These provisions are applicable for either prequalifying precast structural walls for a specific structure or prequalifying a new precast wall type for construction in general. 14.2.4.3.3 Precast structural walls shall be deemed to have a response that is at least equivalent to the response of monolithic structural walls designed in accordance with ACI 318 Sections 21.1 and 21.9, and the corresponding structural walls of the prototype structure shall be deemed acceptable, when all of the conditions in Sections 14.2.4.3.3.1 through 14.2.4.3.3.5 are satisfied. 14.2.4.3.3.1 The prototype structure satisfies all applicable requirements of these provisions and of ACI 318 except Section 21.9. 14.2.4.3.3.2 Tests on wall modules satisfy the conditions in Sections 14.2.4.4 and 14.2.4.9. 14.2.4.3.3.3 The prototype structure is designed using the design procedure substantiated by the testing program. 14.2.4.3.3.4 The prototype structure is designed and analyzed using effective initial properties consistent with those determined in accordance with Section 14.2.4.7.11, and the prototype structure meets the drift limits of these provisions. 14.2.4.3.3.5 The structure as a whole, based on the results of the tests of Section 14.2.4.3.3.2 and analysis, is demonstrated to have adequate global toughness (the ability to retain its structural integrity and support its specified gravity loads) through peak displacements equal to or exceeding the story-drift ratios specified in Section 14.2.4.7.4, 14.2.4.7.5 or 14.2.4.7.6, as appropriate. 14.2.4.4 Design Procedure 14.2.4.4.1 Prior to testing, a design procedure shall be developed for the prototype structure and its walls. That procedure shall account for effects of material non-linearity, including cracking, deformations of members and connections, and reversed cyclic loading. The design procedure shall include the procedures specified in Sections 14.2.4.4.1.1 through 14.2.4.4.1.4 and shall be applicable to all precast structural walls, coupled and uncoupled, of the prototype structure. 14.2.4.4.1.1 Procedures shall be specified for calculating the effective initial stiffness of the precast structural walls, and of coupled structural walls, that are applicable to all the walls of the prototype structure. 14.2.4.4.1.2 Procedures shall be specified for calculating the lateral strength of the precast structural walls, and of coupled structural walls, applicable to all precast walls of the prototype structure. 14.2.4.4.1.3 Procedures shall be specified for designing and detailing the precast structural walls so that they have adequate ductility capacity. These procedures shall cover wall shear strength, sliding shear strength, boundary tie spacing to prevent bar buckling, concrete confinement, reinforcement strain, and any other actions or elements of the wall system that can affect ductility capacity. 14.2.4.4.1.4 Procedures shall be specified for determining that an undesirable mechanism of nonlinear response, such as a story mechanism due to local buckling of the reinforcement or splice failure, or overall instability of the wall, does not occur. 14.2.4.4.2 The design procedure shall be used to design the test modules and shall be documented in the test report. 14.2.4.4.3 The design procedure used to proportion the test specimens shall define the mechanism by which the system resists gravity and earthquake effects and shall establish acceptance values for sustaining that mechanism. Portions of the mechanism that deviate from code requirements shall be contained in the test specimens and shall be tested to determine acceptance values. 14.2.4.5 Test Modules 14.2.4.5.1 At least two modules shall be tested. At least one module shall be tested for each limiting engineering design criteria (shear, axial load and flexure) for each characteristic configuration of precast structural walls, including intersecting structural walls or coupled structural walls. If all the precast walls of the structure have the same configuration and the same limiting engineering design criterion, then two modules shall be tested. Where intersecting precast wall systems are to be used, the response for the two orthogonal directions shall be tested. 14.2.4.5.2 Where the design requires the use of coupling elements, those elements shall be included as part of the test module. 14.2.4.5.3 Modules shall have a scale large enough to represent the complexities and behavior of the real materials and of the load transfer mechanisms in the prototype walls and their coupling elements, if any. Modules shall have a scale not less than one half and shall be full-scale if the validation testing has not been preceded by an extensive analytical and experimental development program in which critical details of connections are tested at full scale. 14.2.4.5.4 The geometry, reinforcing details, and materials properties of the walls, connections, and coupling elements shall be representative of those to be used in the prototype structure. 14.2.4.5.5 Walls shall be at least two panels high unless the prototype structure is one for which a single panel is to be used for the full height of the wall. 14.2.4.5.6 Where precast walls are to be used for bearing wall structures, as defined in ASCE/SEI 7-05, the test modules shall be subject during lateral loading to an axial load stress representative of that anticipated at the base of the wall in the prototype structure. 14.2.4.5.7 The geometry, reinforcing, and details used to connect the precast walls to the foundation shall replicate those to be used in the prototype structure. 14.2.4.5.8 Foundations used to support the test modules shall have geometric characteristics, and shall be reinforced and supported, so that their deformations and cracking do not affect the performance of the modules in a way that would be different than in the prototype structure. 14.2.4.6 Testing Agency. Testing shall be carried out by an independent testing agency approved by the Authority Having Jurisdiction. The testing agency shall perform its work under the supervision of a registered design professional experienced in seismic structural design. 14.2.4.7 Test Method 14.2.4.7.1 Test modules shall be subjected to a sequence of displacement-controlled cycles representative of the drifts expected under earthquake motions for the prototype structure. If the module consists of coupled walls, approximately equal drifts (within 5 percent of each other) shall be applied to the top of each wall and at each floor level. Cycles shall be to predetermined drift ratios as defined in Sections 14.2.4.7.2 through 14.2.4.7.6. 14.2.4.7.2 Three fully reversed cycles shall be applied at each drift ratio. 14.2.4.7.3 The initial drift ratio shall be within the essentially linear elastic response range for the module. See 14.2.4.7.11. Subsequent drift ratios shall be to values not less than 5/4 times, and not more than 3/2 times, the previous drift ratio. 14.2.4.7.4 For uncoupled walls, testing shall continue with gradually increasing drift ratios until the drift ratio in percent equals or exceeds the larger of : (a) 1.5 times the drift ratio corresponding to the design displacement or (b) the following value: (14.2.4-1) where hw = height of entire wall for prototype structure (in inches) and lw = length of entire wall in direction of shear force (in inches). 14.2.4.7.5 For coupled walls, hw/lw in Equation 14.2.4-1 shall be taken as the smallest value of hw/lw for any individual wall of the prototype structure. 14.2.4.7.6 Validation by testing to limiting drift ratios less than those given by Equation 14.2.4-1 shall be acceptable provided testing is conducted in accordance with this document to drift ratios equal or exceeding of those determined for the response to a suite of nonlinear time history analyses conducted in accordance with the 2009 NEHRP Recommended Seismic Provisions for risk-targeted maximum considered earthquake ground motions. 14.2.4.7.7 Actual yield strength of steel reinforcement shall be obtained by testing coupons taken from the same reinforcement batch as used in the test module. Two tests, conforming to the ASTM specifications cited in ACI 318 Section 3.8, shall be made for each reinforcement type and size. Equation Equation 14.2.4.7.8 Actual compressive strength of concrete shall be determined by testing of concrete cylinders cured under the same conditions as the test module and tested at the time of testing the module. Testing shall conform to the applicable requirements of ACI 318 Sections 5.6.1 through 5.6.4. 14.2.4.7.9 Where strength and deformation capacity of coupling devices does not depend on reinforcement tested as required in Section 14.2.4.7.7, the effective yield strength and deformation capacity of coupling devices shall be obtained by testing independent of the module testing. 14.2.4.7.10 Data shall be recorded from all tests such that a quantitative interpretation can be made of the performance of the modules. A continuous record shall be made of test module drift ratio versus applied lateral force, and photographs shall be taken that show the condition of the test module at the peak displacement and after each key testing cycle. 14.2.4.7.11 The effective initial stiffness of the test module shall be calculated based on test cycles to a force between 0.6Ent and 0.9Ent, and using the deformation at the strength of 0.75Ent to establish the stiffness. 14.2.4.8 Test Report 14.2.4.8.1 The test report shall contain sufficient evidence for an independent evaluation of all test procedures, design assumptions, and the performance of the test modules. As a minimum, all of the information required by Sections 14.2.4.8.1.1 through 14.2.4.8.1.11 shall be provided. 14.2.4.8.1.1 A description shall be provided of the design procedure and theory used to predict test module strength, specifically the test module nominal lateral resistance, En, and the test module actual lateral resistance Ent. 14.2.4.8.1.2 Details shall be provided of test module design and construction, including fully dimensioned engineering drawings that show all components of the test specimen. 14.2.4.8.1.3 Details shall be provided of specified material properties used for design, and actual material properties obtained by testing in accordance with Section 14.2.4.7.7. 14.2.4.8.1.4 A description shall be provided of test setup, including fully dimensioned diagrams and photographs. 14.2.4.8.1.5 A description shall be provided of instrumentation, its locations, and its purpose. 14.2.4.8.1.6 A description and graphical presentation shall be provided of applied drift ratio sequence. 14.2.4.8.1.7 A description shall be provided of observed performance, including photographic documentation, of the condition of each test module at key drift ratios including, (as applicable), the ratios corresponding to first flexural cracking or joint opening, first shear cracking, and first crushing of the concrete for both positive and negative loading directions, and any other significant damage events that occur. Photos shall be taken at peak drifts and after the release of load. 14.2.4.8.1.8 A graphical presentation shall be provided of lateral force versus drift ratio response. 14.2.4.8.1.9 A graphical presentation shall be provided of relative energy dissipation ratio versus drift ratio. 14.2.4.8.1.10 A calculation shall be provided of effective initial stiffness for each test module as observed in the test and as determined in accordance with Section 14.2.4.7.11 and a comparison made as to how accurately the design procedure has been able to predict the measured stiffness. The design procedure shall be used to predict the overall structural response and a comparison made as to how accurately that procedure has been able to predict the measured response. 14.2.4.8.1.11 The test date, report date, name of testing agency, report author(s), supervising registered design professional, and test sponsor shall be provided. 14.2.4.9 Test Module Acceptance Criteria 14.2.4.9.1 The test module shall be deemed to have performed satisfactorily when all of the criteria Sections 14.2.4.9.1.1 through 14.2.4.9.1.3 are met for both directions of in-plane response. If any test module fails to pass the validation testing required by these provisions for any test direction, then the wall system has failed the validation testing. 14.2.4.9.1.1 Peak lateral strength obtained shall be at least 0.9Ent and not greater than 1.2 Ent. 14.2.4.9.1.2 In cycling up to the drift level given by Sections 14.2.4.7.4 through 14.2.4.7.6, fracture of reinforcement or coupling elements, or other significant strength degradation, shall not occur. For a given direction, peak lateral strength during any cycle of testing to increasing displacement shall not be less than 0.8 times Emax for that direction. 14.2.4.9.1.3 For cycling at the given drift level for which acceptance is sought in accordance with Section 14.2.4.7.4, 14.2.4.7.5 or 14.2.4.7.6, as applicable, the parameters describing the third complete cycle shall have satisfied the following: 1. The relative energy dissipation ratio shall not be less than 1/8 and 2. The secant stiffness between drift ratios of -1/10 and +1/10 of the maximum applied drift shall not be less than 0.10 times the stiffness for the initial drift ratio specified in Section 14.2.4.7.3. SECTION 14.4.5, MODIFICATIONS TO CHAPTER 1 OF ACI530/ASCE 5/TMS 402 Add the following new sections: 14.4.5.3 Plain (unreinforced) AAC masonry shear walls shall satisfy the requirements of Section 1.14.2.2.6 of ACI 530/ASCE 5/TMS 402. 14.4.5.4 Ordinary reinforced AAC masonry shear walls shall satisfy the requirements of Section 1.14.2.2.8 of ACI 530/ASCE 5/TMS 402. Commentary to Chapter 14 Modifications COMMENTARY TO SECTION 14.1.1 ASCE/SEI 7-05 included two different systems for both eccentrically braced frames (EBF) and buckling restrained braced frames (BRBF). The primary distinction between these two systems was whether or not there were moment resisting beamcolumn connections within the braced bays. However, testing at the University of California at Berkeley (Uriz and Mahin, 2004) has indicated designs that do not properly account for the stiffness and distribution of forces in braced frame connections may be subject to undesirable performance. This modification to ASCE/SEI 7-05 consolidates the EBF and BRBF building frame systems into a single designation with proper consideration of the beam-column connection demands. This modification to ASCE/SEI 7-05 and the related changes to AISC 341-05 Sections 15.7 and 16.7 also allow the engineer either to: 1. Provide a fully restrained moment connection meeting the requirements for ordinary moment connections in AISC 341- 05 and thereby directly providing a load path to resist the connection force and deformation demands or 2. Provide a connection with the ability to accommodate the potential rotation demands. An example of this would be a configuration tested at Lehigh University (Figure 1 of Fahnestock, et. al. 2006) that effectively formed a pinned condition in the beam just beyond the beam-column-brace connection. COMMENTARY TO SECTION 14.1.4 C14.1.4 Cold-Formed Steel. This section adopts three standards by direct reference: AISI NAS, North American Specification for the Design of Cold-Formed Steel Structural Members, AISI S110, Standard for Seismic Design of Cold- Formed Steel Structural Systems – Special Bolted Moment Frames, and ASCE/SEI 8, Specification for the Design of Cold Formed Stainless Steel Structural Members. Each document has specific limits of applicability. AISI NAS applies to the design of structural members that are coldformed to shape from carbon or low-alloy steel sheet, strip, plate or bar not more than one-inch in thickness (AISI NAS, Section A1.1). Building on the requirements of AISI NAS, AISI S110 has additional special seismic design provisions for a newly designated seismic force resisting system entitled “cold-formed steel – special bolted moment frame (CFS-SBMF).” Finally, ASCE 8 governs the design of structural members that are cold-formed to shape from annealed and cold-rolled sheet, strip, plate, or flat bar stainless steels (ASCE 8, Section 1.1.1). All three documents focus on load-carrying members in buildings; however, allowances are made for applications in nonbuilding structures, if dynamic effects are appropriately considered. Within AISI NAS and ASCE 8, there are requirements on the general provisions for the applicable types of steel; design of elements, members, structural assemblies, connections and joints; and mandatory testing. In addition, AISI NAS contains a chapter on the design of cold-formed steel structural members and connections undergoing cyclic loading. Both standards contain extensive commentaries for the benefit of the user. C14.1.4.1.1 CFS-SBMF need to use the same-size beams and same-size columns throughout. In addition, the system needs to engage all primary columns, which support the roof or floor above, and those columns need to be supported on a level floor or foundation. C14.1.4.1.2 These modifications were made for consistency with the test database. C14.1.4.1.3 To be consistent with the test database (Uang and Sato, 2007), the limitations on both beam depth, steel grade, and surface treatment are added in Section D1.2.1 of AISI S110. C14.1.4.1.4 To be consistent with the test database (Uang and Sato, 2007), the limitations on column depth, steel grade, and surface treatment are added in Section D1.2.2 of AISI S110. The width-thickness ratio was reduced based upon further review of the test specimens. C14.1.4.1.5 AISI S110 is intended primarily for industrial platforms; however, the standard is not limited to these nonbuilding structures and does not prohibit architectural attachments (such as partition walls). Therefore, the 0.05h drift limit in Section D1.3 of AISI S110 has been reduced to 0.03h to more closely align with the 0.025h drift limit of ASCE/SEI 7. The sentence, “In no case shall the design story drift exceed 0.05h.” was added to ensure an absolute upper bound on the drift limit. C14.1.4.2 Light-Framed Cold-Formed Construction. This subsection of cold-formed steel relates to light-framed construction, which is defined as a method of construction where the structural assemblies are formed primarily by a system of repetitive wood or cold-formed steel framing members or subassemblies of these members (ASCE/SEI 7, Section 11.2). Not only does this subsection repeat the direct adoptions of AISI NAS and ASCE 8, but it also allows the user to choose from an additional suite of standards that address different aspects of construction, including the following: 1. AISI GP, Standard for Cold-Formed Steel Framing – General Provisions, applies to the design, construction, and installation of structural and non-structural cold-formed steel framing members where the specified minimum base metal thickness is between 18 mils and 118 mils (AISI GP, Section A1). 2. AISI WSD, Standard for Cold-Formed Steel Framing – Wall Stud Design, applies to the design and installation of coldformed steel studs for both structural and nonstructural walls in buildings (AISI WSD, Section A1). COMMENTARY TO SECTION 14.2 C14.2 CONCRETE The section adopts by reference ACI 318 for structural concrete design and construction. In addition, modifications to ACI 318 are made that are needed to coordinate the provisions of that material design standard with the provisions of ASCE/SEI 7. Work is ongoing to better coordinate the provisions of the two documents (ACI 318 and ASCE/SEI 7) such that the provisions in Section 14.2 will be significantly reduced in future editions of ASCE/SEI 7. C14.2.2.2 ACI 318 Section 7.10. ACI 318 Section 7.10.5.6 prescribes reinforcement details for ties in compression members. Those details are appropriate for SDC A and B structures. This modification prescribes additional details for ties around anchor bolts in structures assigned to SDC C through F. A wall pier is recognized as a separate category of structural element in this document but not ACI 318. C14.2.2.3 Scope. This provision describes how the ACI 318 provisions should be interpreted for consistency with the ASCE/SEI 7 provisions. C14.2.2.4 Intermediate Precast Structural Walls. ACI 318 Section 21.4 imposes requirements on precast walls for moderate seismic risk applications. Ductile behavior is to be ensured by yielding of the steel elements or reinforcement between panels or this provision requires the designer to determine the deformation in the connection corresponding to the earthquake design displacement, and then to check from experimental data that the connection type used can accommodate that deformation without significant strength degradation. The wall pier requirements of Section 21.4.5 are patterned after the same requirements of Section 14.2.2.4 for wall piers that are part of structures in high seismic design categories. The 2006 Edition of the International Building Code restricts yielding to steel reinforcement only because of concern that steel elements in the body of a connection could fracture due to inelastic strain demands. Several steel element connections have been tested under simulated seismic loading and the adequacy of their loaddeformation characteristics and strain capacity have been demonstrated (Schultz and Magana, 1996). One such connection was used in the five-story building test that was part of the PRESSS Phase 3 research. The connection was used to provide damping and energy dissipation, and demonstrated a very large strain capacity (Nakaki et al., 2001). Since then, several other steel element connections have been developed that can achieve similar results (Banks and Stanton), (Nakaki et al.). In view of these results, it is appropriate to allow yielding in steel elements that have been shown experimentally to have adequate strain capacity to maintain at least 80 percent of their yield force through the full design displacement of the structure. C14.2.2.5 Wall Piers and Wall Segments. Wall piers are typically segments between openings in walls that are thin in the direction normal to the horizontal length of the wall. In current practice these elements are often not regarded as columns or as part of the structural walls. If not properly reinforced these elements are vulnerable to shear failure and that failure prevents the wall from developing the assumed flexural hinging. Section 21.9.10 is written to reduce the likelihood of a shear failure. Wall segments with a horizontal length-to-thickness ratio less than 2.5 are required to be designed as columns in compliance with Section 21.9 if they are utilized as part of the lateral-force-resisting system, even though the shortest cross-sectional dimension may be less than 12 in. in violation of Section 21.6.1.1. Such wall segments may be designed to comply with Section 21.13 if they are not utilized as part of the lateral-force-resisting system. Wall segments with a horizontal length-to-thickness ratio larger than or equal to 2.5, which do not meet the definition of wall piers (Section 14.2.2.2), must be designed as special structural walls or as portions of special structural walls in full compliance with Section 21.9 or 21.10. C14.2.2.7 Foundations. The intention is that there should be no conflicts between the provisions of ACI 318 Section 21.12 and ASCE/SEI 7-05 Section 12.1.5, 12.13, or 14.2. However, the additional detailing requirements for concrete piles of Section 14.2.3 can result in conflicts with ACI 318 provisions if the pile in not fully embedded in the soil. C14.2.2.8 Detailed Plain Concrete Walls. Design requirements for plain masonry walls have existed for many years and the corresponding type of concrete construction is the plain concrete wall. To allow the use of such walls as the lateral-forceresisting system in SDC A and B, this provision requires such walls to contain at least the minimal reinforcement specified in Section 22.6.7.2. C14.2.2.9 Strength Requirements for Anchors. ACI 318 requires laboratory testing to establish the strength of anchor bolts greater than 2 in. in diameter or exceeding 25 in. in tensile embedment depth. This modification makes the ACI 318 equation giving the basic concrete breakout strength of a single anchor in tension in cracked concrete applicable irrespective of the anchor bolt diameter and tensile embedment depth. Korean Power Engineering (KPE) has made tension tests on anchors with diameters up to 4.25 in. and embedment depths up to 45 in. and found that the diameter and embedment depth limits of ACI 318 Section D.4.2.2 for the design procedure for anchors in tension (Section D.5.2) can be eliminated. KPE also has conducted shear tests on anchors with diameters up to 3.0 in. and embedment depths as large as 30 in. and found no effect of the embedment depth on shear strength. However, the diameter tests showed that the basic shear breakout strength Equation D-24 needed some modification for the complete elimination of the 2 in. limit to be fully appropriate. Analytical work performed at the University of Stuttgart supports the need for some modification to Equation D-24. Changes consistent with the Korean and Stuttgart findings have already been made to the FIB Design Guide for anchors. COMMENTARY TO SECTION 14.2.3 C14.2.3 Additional Detailing Requirements for Concrete Piles. Chapter 20 of the PCI Bridge Design Manual provides detailed information on the structural design of piles and on pile to cap connections for precast prestressed concrete piles. ACI 318 does not contain provisions governing the design and installation of portions of concrete piles, drilled piers, and caissons embedded in ground except for SDC D, E and F structures. C14.2.3.1.2 Reinforcement for Uncased Concrete Piles (SDC C). The transverse reinforcing requirements in the potential plastic hinge zone of uncased concrete piles in Seismic Design Category C is a selective composite of two ACI 318 requirements. In the potential plastic hinge region of an intermediate moment-resisting concrete frame column, the transverse reinforcement spacing is restricted to the least of: (a) 8 times the diameter of the smallest longitudinal bar, (b) 24 times the diameter of the tie bar, (c) one-half the smallest cross-sectional dimension of the column, and (d) 12 in. Outside of the potential plastic hinge region of a special moment-resisting frame column, the transverse reinforcement spacing is restricted to the smaller of 6 times the diameter of the longitudinal column bars and 6 in. C14.2.3.1.5 Reinforcement for Precast Nonprestressed Concrete Piles (SDC C). Transverse reinforcement requirements in and outside of the plastic hinge zone of precast nonprestressed piles are clarified. The transverse reinforcement requirement in the potential plastic hinge zone is a composite of two ACI 318 requirements (see Section C14.2.3.1.2). Outside of the potential plastic hinge region, the transverse reinforcement spacing is restricted to sixteen (16) times the longitudinal bar diameter. This should permit the longitudinal bars to reach compression yield before buckling. The maximum 8-in. tie spacing comes from current building code provisions for precast concrete piles. C14.2.3.1.6 Reinforcement for Precast Prestressed Piles (SDC C). The transverse and longitudinal reinforcing requirements given in ACI 318, Chapter 21, were never intended for slender precast prestressed concrete elements and will result in unbuildable piles. These requirements are based on the Recommended Practice for Design, Manufacture and Installation of Prestressed Concrete Piling (PCI Committee on Prestressed Concrete Piling, 1993). Equation 14.2.4-1, originally from ACI 318, has always been intended to be a lower-bound spiral reinforcement ratio for larger diameter columns. It is independent of the member section properties and can therefore be applied to large or small diameter piles. For cast-in-place concrete piles and precast prestressed concrete piles, the resulting spiral reinforcing ratios from this formula are considered to be sufficient to provide moderate ductility capacities (Fanous et al., 2007). Full confinement per Equation 14.2.4-1 is required for the upper 20 feet of the pile length where curvatures are large. The amount is relaxed by 50 percent outside of that length in view of lower curvatures and in consideration of confinement provided by the soil. C14.2.3.2.3 Reinforcement for Uncased Concrete Piles (SDC D through F). The reinforcement requirements for uncased concrete piles are taken from the 2006 IBC requirements, and should be adequate to provide ductility in the potential plastic hinge zones (Fanous et al., 2007). C14.2.3.2.5 Reinforcement for Precast Concrete Piles (SDC D through F). The transverse reinforcement requirements for precast nonprestressed concrete piles are taken from the 2006 IBC requirements and are should be adequate to provide ductility in the potential plastic hinge zones (Fanous et al., 2007). C14.2.3.2.6 Reinforcement for Precast-Prestressed Piles (SDC D through F). The reduced amounts of transverse reinforcement specified in this provision compared to those required for column members in ACI 318 are justified by the results of the study by Fanous et al., 2007. The last paragraph of the provision provides minimum transverse reinforcement requirements outside of the zone of prescribed ductile detailing. COMMENTARY TO SECTION 14.2.4 C14.2.4 Acceptance Criteria for Special Precast Structural Walls Based on Validation Testing C14.2.4.1 Notation. Symbols additional to those in ACI 318 Chapter 2 are defined: Ah = area of hysteresis loop. E1,E2 = peak lateral resistance for positive and negative loading, respectively, for third cycle of loading sequence. f1 = live load factor defined in Section 14.2.4.2.3. hw = height of column of test module, in. or mm. K, K’ = initial stiffness for positive and negative loading, respectively, for first cycle. .1,.2 = drift ratios at peak lateral resistance for positive and negative loading, respectively, for third cycle of loading sequence. .1',.2' = drift ratios for zero lateral load for unloading at stiffness K, K’ from peak positive and negative lateral resistance, respectively, for third cycle of loading sequence. . = lateral displacement, in. or mm. See Figures. C14.2.4.2.2-1, C14.2.4.2.2-2, and C14.2.4.2.2-3. .a = allowable story drift, in. or mm. See Table 12.12-1 of ASCE/SEI 7-05. C14.2.4.2 Definitions C14.2.4.2.1 Coupling elements. Coupling elements are connections provided at specific intervals along the vertical boundaries of adjacent structural walls. Coupled structural walls are stiffer and stronger than the same walls acting independently. For cast-in-place construction effective coupling elements are typically coupling beams having small span-todepth ratios. The inelastic behavior of such beams is normally controlled by their shear strength. For precast construction, effective coupling elements can be precast beams connected to the adjacent structural walls either by post-tensioning, ductile mechanical devices, or grouted-in-place reinforcing bars. The resultant coupled construction can be either emulative of castin- place construction or non-emulative (jointed). However, for precast construction coupling beams can also be omitted and mechanical devices used to connect directly the vertical boundaries of adjacent structural walls. C14.2.4.2.2 Drift ratio. The definition of the drift ratio, ., is illustrated in Figure C14.2.4.2.2-1 for a three panel wall module. The position of the module at the start of testing, with only its self-weight acting, is indicated by broken lines. The module is set on a horizontal foundation support that is centered at A and is acted on by a lateral force H applied at the top of the wall. The self-weight of the wall is distributed uniformly to the foundation support. However, under lateral loading, that self-weight and any axial gravity load acting at the top of the wall cause overturning moments on the wall that are additional to the overturning moment Hhw and can affect deformations. The chord AB of the centroidal axis of the wall is the vertical reference line for drift measurements. For acceptance testing a lateral force H is applied to the wall through the pin at B. Depending on the geometric and reinforcement characteristics of the module that force can result in the module taking up any one, or a combination, of the deformed shapes indicated by solid lines in Figures C14.2.4.2.2-1, C14.2.4.2.2-2 and C14.2.4.2.2-3. Figure C14.2.4.2.2-2 illustrates several possible components of the displacement . for a wall that is effectively solid while Figure C14.2.4.2.2-3 illustrates two possibly undesirable components of the displacement .. Regardless of the mode of deformation of the wall, the lateral force causes the wall at B to displace horizontally by an amount .. The drift ratio is the angular rotation of the wall chord with respect to the vertical and for the setup shown equals . / hw where hw is the wall height and is equal to the distance between the foundation support at A and the load point at B. Where prestressing steel is used in wall members, the stress fps in the reinforcement at the nominal and the probable lateral resistance shall be calculated in accordance with ACI 318 Section 18.7. C14.2.4.2.3 Global toughness. These provisions describe acceptance criteria for special precast structural walls based on validation testing. The requirements of Section 21.1.1.8 of ACI 318 concerning toughness cover both to the energy dissipation of the wall system which, for monolithic construction, is affected primarily by local plastic hinging behavior and the toughness of the prototype structure as a whole. The latter is termed “global toughness” in these provisions and is a condition that does not apply to the walls alone. That global toughness requirement can be satisfied only though analysis of the performance of the prototype structure as a whole when the walls perform to the criteria specified in these provisions. The required gravity load for global toughness evaluations is the value given by these provisions. For conformity with Section 9.2.1 of ACI 318-08, UBC 1997, IBC 2006 and NFPA 5000, the required gravity load is 1.2D + f1L where the seismic force is additive to gravity forces and 0.9D where the seismic force counteracts gravity forces. D is the effect of dead loads, L is the effect of live loads, and f1 is a factor equal to 0.5 except for garages, areas occupied as places of public assembly, and all areas where the live load is greater than 100 psf (4.79 kN/m2) where f1 equals 1.0. C14.2.4.2.5 Relative energy dissipation ratio. This concept is illustrated in Figure C14.2.4.2.2-1 for the third loading cycle to the limiting drift ratio required by Section 14.2.4.7.4, 14.2.4.7.5 or 14.2.4.7.6, as appropriate. Figure 14.2.4.2.2-2 illustrates several possible components of the displacement . for a wall that is effectively solid while Figure C14.2.4.2.2-3 illustrates two possibly undesirable components of the displacement .. Regardless of the mode of deformation of the wall, the lateral force causes the wall at B to displace horizontally by an amount .. The drift ratio is the angular rotation of the wall chord with respect to the vertical and for the setup shown equals . / hw where hw is the wall height and is equal to the distance between the foundation support at A and the load point at B. Where prestressing steel is used in wall members, the stress fps in the reinforcement at the nominal and the probable lateral resistance shall be calculated in accordance with Section 18.7 of ACI 318. C14.2.4.2.3 Global toughness. These provisions describe acceptance criteria for special precast structural walls based on validation testing. The requirements of ACI 318 Section 21.1.1.8 concerning toughness cover both to the energy dissipation of the wall system which, for monolithic construction, is affected primarily by local plastic hinging behavior and the toughness of the prototype structure as a whole. The latter is termed “global toughness” in these provisions and is a condition that does not apply to the walls alone. That global toughness requirement can be satisfied only though analysis of the performance of the prototype structure as a whole when the walls perform to the criteria specified in these provisions. The required gravity load for global toughness evaluations is the value given by these provisions. For conformity with Section 9.2.1 of ACI 318-08, UBC 1997, IBC 2006 and NFPA 5000, the required gravity load is 1.2D + f1L where the seismic force is additive to gravity forces and 0.9D where the seismic force counteracts gravity forces. D is the effect of dead loads, L is the effect of live loads, and f1 is a factor equal to 0.5 except for garages, areas occupied as places of public assembly, and all areas where the live load is greater than 100 psf (4.79 kN/m2) where f1 equals 1.0. C14.2.4.2.5 Relative energy dissipation ratio. This concept is illustrated in Figure C14.2.4.2.5 for the third loading cycle to the limiting drift ratio required by Section 14.2.4.7.4, 14.2.4.7.5 or 14.2.4.7.6, as appropriate. For Figure C14.2.4.2.5, it is assumed that the test module has exhibited different initial stiffnesses, K and K’, for positive and negative lateral forces and that the peak lateral resistances for the third cycle for the positive and negative loading directions, E1 and E2, also differ. The area of the hysteresis loop for the third cycle, Ah, is hatched. The circumscribing figure consists of two parallelograms, ABCD and DFGA. The slopes of the lines AB and DC are the same as the initial stiffness, K, for positive loading and the slopes of the lines DF and GA are the same as the initial stiffness, K', for negative loading. The relative energy dissipation ratio concept is similar to the equivalent damping concept used in Section 17.8.3 of the ASCE/SEI 7-05 for required tests of seismic isolation systems. Figure C14.2.4.2.2-1 Showing Definition of drift ratio .. Figure C14.2.4.2.2-3 Showing Undesirable deformations along horizontal joints: (a) excessive gap opening between panels and (b) shear slip. Figure C14.2.4.2.2-2 Showing Typical wall deformation components. Figure C14.2.4.2.2-1 Definition of drift ratio .. Figure C14.2.4.2.2-2 Typical wall deformation components. (c) Deformation due to shear (b) Deformation due to flexure (d) Deformation due to extension of reinforcement at foundation to wall interface (a) Wall and loading Figure C14.2.4.2.2-3 Undesirable deformations along horizontal joints: (a) excessive gap opening between panels and (b) shear slip. Figure C14.2.4.2.5 Showing Relative energy dissipation ratio. For a given cycle the relative energy dissipation ratio, ß, is the area, Ah, inside the lateral force-drift ratio loop for the module, divided by the area of the effective circumscribing parallelograms ABCD and DFGA. The areas of the parallelograms equal the sum of the absolute values of the lateral force strengths, E1 and E2, at the drift ratios .1 and .2 multiplied by the sum of the absolute values for the drift ratios .1' and .2'. C14.2.4.3 Scope and general requirements. While only ACI Committee 318 can determine the requirements necessary for precast walls to meet the provisions of ACI 318 Section 21.1.1.8, ACI 318 Section 1.4 already permits the building official to accept wall systems, other than those explicitly covered by ACI 318 Chapter 21, provided specific tests, load factors, deflection limits, construction procedures and other pertinent requirements have been established for acceptance of such systems consistent with the intent of the code. The purpose of these provisions is to provide a framework that establishes the specific tests, load factors, deflection limits and other pertinent requirements appropriate for acceptance, for regions of high seismic risk or for structures assigned to high seismic performance or design categories, of precast wall systems, including coupled wall systems, not satisfying all the requirements of ACI 318 Chapter 21. For regions of moderate seismic risk or for structures assigned to intermediate seismic performance or design categories, less stringent provisions than those specified here are appropriate. These provisions assume that the precast wall system to be tested has details differing from those prescribed by ACI 318 Section 21.9 for conventional monolithic reinforced concrete construction. Such walls may, for example, involve the use of precast elements, precast prestressed elements, post-tensioned reinforcement, or combinations of those elements and reinforcement. Figure C14.2.4.2.5 Relative energy dissipation ratio. For monolithic reinforced concrete walls a fundamental design requirement of ACI 318 Chapter 21 is that walls with hw/lw exceeding1.0 be proportioned so that their inelastic response is dominated by flexural action on a critical section located near the base of the wall. That fundamental requirement is retained in these provisions. The reason is that tests on modules, as envisioned in these provisions, cannot be extrapolated with confidence to the performance of panelized walls of proportions differing from those tested for the development of ACI 318 Chapter 21 if the shear-slip displacement pattern of Figure C14.2.4.2.2.3, or the shear deformation response of Figure C14.2.4.2.2.2, governs the response developed in the test on the module. Two other fundamental requirements of ACI 318 Chapter 21 are for ties around heavily strained boundary element reinforcement and the provision of minimum amounts of uniformly distributed horizontal and vertical reinforcement in the web of the wall. Ties around boundary element reinforcement to inhibit its buckling in compression are required where the strain in the extreme compression fiber is expected to exceed some critical value. Minimum amounts of uniformly distributed horizontal and vertical reinforcement over the height and length of the wall are required to restrain the opening of inclined cracks and allow the development of the drift ratios specified in Sections 14.2.4.7.4, 14.2.4.7.5 and 14.2.4.7.6. Deviations from those tie and distributed reinforcement requirements are possible only if a theory is developed that can substantiate reasons for such deviations and that theory is tested as part of the validation testing. C14.2.4.3.1. These provisions are not intended for use with existing construction or for use with walls that are designed to conform to all the requirements of ACI 318 Section 21.9. The criteria of these provisions are more stringent than those for walls designed to ACI 318 Section 21.9. Some walls designed to Section 21.9, and having low height to length ratios, may not meet the drift ratio limits of Equation 14.2.4-1 because their behavior may be governed by shear deformations. The height to length ratio of 0.5 is the least value for which Equation 14.2.4-1 is applicable. C14.2.4.3.3 For acceptance, the results of the tests on each module must satisfy the acceptance criteria of Section 14.2.4.9. In particular, the relative energy dissipation ratio calculated from the measured results for the third cycle between the specified limiting drift ratios must equal or exceed 1/8. For uncoupled walls, relative energy dissipation ratios increase as the drift ratio increases. Tests on slender monolithic walls have shown relative energy dissipation ratios, derived from rotations at the base of the wall, of about 40-45 percent at large drifts. The same result has been reported even where there has been a significant opening in the web of the wall on the compression side. For 0.020 drift ratios and walls with height to length ratios of 4, relative energy dissipation ratios have been computed as 30, 18, 12, and 6 percent, for monolithic reinforced concrete, hybrid reinforced/post-tensioned prestressed concrete with equal flexural strengths provided by the prestressed and deformed bar reinforcement, hybrid reinforced/post-tensioned prestressed concrete with 25 percent of the flexural strength provided by deformed bar reinforcement and 75 percent by the prestressed reinforcement, and post-tensioned prestressed concrete special structural walls, respectively. Thus, for slender precast uncoupled walls of emulative or non-emulative design it is to be anticipated that at least 35 percent of the flexural capacity at the base of the wall needs to be provided by deformed bar reinforcement if the requirement of a relative energy dissipation ratio of 1/8 is to be achieved. However, if more than about 40 percent of the flexural capacity at the base of the wall is provided by deformed bar reinforcement, then the self-centering capability of the wall following a major event is lost and that is one of the prime advantages gained with the use of post-tensioning. For squat walls with height to length ratios between 0.35 and 0.69 the relative energy dissipation has been reported as remaining constant at 23 percent for drifts between that for first diagonal cracking and that for a postpeak capacity of 80 percent of the peak capacity. Thus, regardless of whether the behavior of a wall is controlled by shear or flexural deformations a minimum relative energy dissipation ratio of 1/8 is a realistic requirement. For coupled wall systems, theoretical studies and tests have demonstrated that the 1/8 relative energy dissipation ratio can be achieved by using central post-tensioning only in the walls and appropriate energy dissipating coupling devices connecting adjacent vertical wall boundaries. C14.2.4.3.3.4. The ASCE/SEI 7-05 allowable story drift limits are the basis for the drift limits of IBC 2006 and NFPA 5000. Allowable story drifts, .a, are specified in Table 1617.3 of IBC 2006 and likely values are discussed in the Commentary to Section 14.2.4.7.4. The limiting initial drift ratio consistent with .a equals .a/fCdhw, where f is the strength reduction factor appropriate to the condition, flexure or shear, that controls the design of the test module. For example, for .a/hw equal to 0.015, the required deflection amplification factor Cd of 5, and f equal to 0.9, the limiting initial drift ratio, corresponding to B in Figure C14.2.4.9.1, is 0.0033. The use of a f value is necessary because the allowable story drifts of the IBC are for the design seismic load effect, E, while the limiting initial drift ratio is at the nominal strength, En , which must be greater than E/f. The load-deformation relationship of a wall becomes significantly non-linear before the applied load reaches Ent. While the load at which that non-linearity becomes marked depends on the structural characteristics of the wall, the response of most walls remains linear up to about 75 percent of Ent. C14.2.4.3.3.5. The criteria of Section 14.2.4.9 are for the test module. In contrast, the criterion of Section 14.2.4.3.3.5 is for the structural system as a whole and can be satisfied only by the philosophy used for the design and analysis of the building as a whole. The criterion adopted here is similar to that described in the last paragraph of R21.1.1 of ACI 318 and the intent is that test results and analyses demonstrate that the structure, after cycling three times through both positive and negative values of the limiting drift ratio specified in Section 14.2.4.7.4, 14.2.4.7.5 or 14.2.4.7.6, as appropriate, is still capable of supporting the gravity load specified as acting on it during the earthquake. Figure C14.2.4.9.1 Quantities used in evaluating acceptance criteria. Figure C14.2.4.9.1 Quantities used in evaluating acceptance criteria. C14.2.4.4 Design Procedure. C14.2.4.4.1. The test program specified in these provisions is intended to verify an existing design procedure for precast structural walls for a specific structure or for prequalifying a generic type of special precast wall system for construction in general. The test program is not for the purpose of creating basic information on the strength and deformation properties of such systems for design purposes. Thus, the test modules should not fail during the validation testing, a result that is the opposite of what is usually necessary during testing in the development phase for a new or revised design procedure. For a generic precast wall system to be accepted based on these provisions, a rational design procedure is to have been developed prior to this validation testing. The design procedure is to be based on a rational consideration of material properties and force transfer mechanisms, and its development will usually require preliminary and possibly extensive physical testing that is not part of the validation testing. Because special wall systems are likely to respond inelastically during design-level ground shaking, the design procedure must consider wall configuration, equilibrium of forces, compatibility of deformations, the magnitudes of the lateral drifts, reversed cyclic displacements, the relative values of each limiting engineering design criteria (shear, flexure and axial load) and use appropriate constitutive laws for materials that include considerations of effects of cracking, loading reversals and inelasticity. The effective initial stiffness of the structural walls is important for calculating the fundamental period of the prototype structure. The procedure used to determine the effective initial stiffness of the walls is to be verified from the validation test results as described in Section 14.2.4.7.11. Provisions Sections 14.2.4.4.1.1 through 14.2.4.4.1.3 state the minimum procedures to be specified in the design procedure prior to the start of testing. The Authority Having Jurisdiction may require that more details be provided in the design procedure than those of Sections 14.2.4.4.1.1 through 14.2.4.4.1.3 prior to the start of testing. C14.2.4.4.2. The justification for the small number of test modules, specified in Section 14.2.4.5.1 is that a previously developed rational design procedure is being validated by the test results. Thus, the test modules for the experimental program must be designed using the procedure intended for the prototype wall system and strengths must be predicted for the test modules before the validation testing is started. Figure C14.2.4.5.1 (a) Showing Coupled wall test module with coupling beams and (b) Showing Coupled wall test module with vertical mechanical couplers. C14.2.4.5 Test Modules. C14.2.4.5.1. One module must be tested for each limiting engineering design criterion, such as shear, or axial load and flexure, for each characteristic configuration of walls. Thus, in accordance with Section 14.2.4.4.3 if the test on the module results in a maximum shear stress of 3vfc’ then the maximum shear stress that can be used in the prototype is that same value. Each characteristic in-plane configuration of walls, or coupled walls, in the prototype structure must also be tested. Thus, as a minimum for one-way structural walls, two modules with the configuration shown in Figure C14.2.4.2.2-1, and, for one way coupled walls, two modules with the configuration shown in either Figure C14.2.4.5.1(a) or in Figure C14.2.4.5.1(b), must be tested. In addition, if intersecting wall systems are to be used then the response of the wall systems for the two orthogonal directions needs to be tested. For two-way wall systems and coupled wall-frame systems, testing of configurations other than those shown in Figures C14.2.4.2.2-1 and C14.2.4.5.1 may be appropriate when it is difficult to realistically model the likely dominant earthquake deformations using orthogonal direction testing only. This provision should not be interpreted as implying that only two tests will need to be made to qualify a generic system. During the development of that system it is likely that several more tests will have been made, resulting in progressive refinements of the mathematical model used to describe the likely performance of the generic structural wall system and its construction details. Consequently, only one test of each module type for each limiting engineering design condition, at a specified minimum scale and subjected to specific loading actions, may be required to validate the system. Further, as stated in Section 14.2.4.9.1, if any one of those modules for the generic wall system fails to pass the validation testing required by these provisions, then the generic wall system has failed the validation testing In most prototype structures, a slab is usually attached to the wall and, as demonstrated by the results of the PRESSS building test, the manner in which the slab is connected to the wall needs to be carefully considered. The connection needs to be adequate to allow the development of story drifts equal to those anticipated in these provisions. However, in conformity with common practice for the sub-assemblage tests used to develop the provisions of Chapter 21of ACI 318, there is no requirement for a slab to be attached to the wall of the test module. The effect of the presence of the slab should be examined in the development program that precedes the validation testing. C14.2.4.5.3. Test modules need not be as large as the corresponding walls in the prototype structure. The scale of the test modules, however, must be large enough to capture all the complexities associated with the materials of the prototype wall, its geometry and reinforcing details, load transfer mechanisms, and joint locations. For modules involving the use of precast elements, for example, scale effects for load transfer through mechanical connections should be of particular concern. The issue of the scale necessary to capture fully the effects of details on the behavior of the prototype should be examined in the development program that precedes the validation testing. drift angle, . angle, . drift angle, . drift centrally post-tensioned coupling beams grouted deformed top bars deflected configuration central unbonded post-tensioning central unbonded post-tensioning undeflected position relative vertical deflection drift drift undeflected position mechanical coupling devices deflected configuration Figure C14.2.4.5.1 (a) Coupled wall test module with coupling beams; (b) Coupled wall test module with vertical mechanical couplers. C14.2.4.5.4. It is to be expected that for a given generic precast wall structure, such as an unbonded centrally post-tensioned wall constructed using multiple precast or precast pretensioned concrete wall panels, validation testing programs will initially use specific values for the specified strength of the concrete and reinforcement in the walls, the layout of the connections between panels, the location of the post-tensioning, the location of the panel joints, and the design stresses in the wall. Pending the development of an industry standard for the design of such walls, similar to the standard for special hybrid moment frames, specified concrete strengths, connection layouts, post-tensioning amounts and locations, etc., used for such walls will need to be limited to the values and layouts used in the validation testing programs. C14.2.4.5.5. For walls constructed using precast or precast/prestressed panels and designed using non-emulative methods, the response under lateral load can change significantly with joint opening (Figure C14.2.4.2.2-2d and Figure C14.2.4.2.2- 3a). The number of panels used to construct a wall depends on wall height and design philosophy. If, in the prototype structure, there is a possibility of horizontal joint opening under lateral loading at a location other than the base of the wall, then the consequences of that possibility need to be considered in the development and validation test programs. Joint opening at locations other than the base can be prevented through the use of capacity design procedures. C14.2.4.5.6. The significance of the magnitude of the gravity load that acts simultaneously with the lateral load needs to be addressed during the validation testing if the development program suggests that effect is significant. C14.2.4.5.7. Details of the connection of walls to the foundation are critical, particularly for non-emulative wall designs. The deformations that occur at the base of the wall due to plastic hinging or extension of the reinforcing bars or posttensioning steel crossing the wall to foundation interface, (Figure C14.2.4.2.2-2d), are in part determined by details of the anchorage and the bonding of those reinforcements on either side of the interface. Grout will be normally used to bed panels on the foundation and the characteristics of that grout in terms of materials, strength and thickness, can have a large effect on wall performance. The typical grout pad with a thickness of 1 inch (25 mm) or less can be expected to provide a coefficient of friction of about 0.6 under reversed loadings. Pads with greater thickness and without fiber reinforcement exhibit lesser coefficients of friction. Adequate frictional resistance is essential to preventing undesirable shear-slip deformations of the type shown in Figure C14.2.4.2.2.3(b). C14.2.4.5.8. The geometry of the foundations need not duplicate that used in the prototype structure. However, the geometric characteristics of the foundations (width, depth and length) need to be large enough that they do not influence the behavior of the test module. C14.2.4.6 Testing Agency. In accordance with the spirit of the requirements of Sections 1.3.5 and 1.4 of ACI 318, it is important that testing be carried out by a recognized independent testing agency, approved by the agency having jurisdiction and that the testing and reporting be supervised by a registered design professional familiar with the proposed design procedure and experienced in testing and seismic structural design. C14.2.4.7 Test Method. The test sequence is expressed in terms of drift ratio, and the initial ratio is related to the likely range of linear elastic response for the module. That approach, rather than testing at specific drift ratios of 0.005, 0.010, etc., is specified because, for modules involving prestressed concrete, the likely range of elastic behavior varies with the prestress level. An example of the test sequence specified in Sections 14.2.4.7.2 through 14.2.4.7.6 is illustrated in Figure C14.2.4.7. The sequence is intended to ensure that displacements are increased gradually in steps that are neither too large nor too small. If steps are too large, the drift capacity of the system may not be determined with sufficient accuracy. If the steps are too small, the system may be unrealistically softened by loading repetitions, resulting in artificially low maximum lateral resistances and artificially high maximum drifts. Also, when steps are too small, the rate of change of energy stored in the system may be too small compared with the change occurring during a major event. Results, using such small steps, can mask undesirable brittle failure modes that might occur in the inelastic response range during a major event. Because significant diagonal cracking is to be expected in the inelastic range in the web of walls, and in particular in squat walls, the pattern of increasing drifts used in the test sequence can markedly affect diagonal crack response in the post-peak range of behavior. The drift capacity of a building in a major event is not a single quantity, but depends on how that event shakes the structure. In the forward near field, a single pulse may determine the maximum drift demand, in which case a single large drift demand cycle for the test module would give the best estimation of the drift capacity. More often, however, many small cycles precede the main shock and that is the scenario represented by the specified loading. Figure C14.2.4.7 Showing Example of specified test sequence. There is no requirement for an axial load to be applied to the wall simultaneously with the application of the lateral displacements. In many cases it will be conservative not to apply axial load because, in general, the shear capacity of the wall and the resistance to slip at the base of the wall increase as the axial load on the wall increases. However, as the height of the wall increases and the limiting drift utilized in the design of the wall increases, the likelihood of extreme fiber crushing in compression at maximum drift increases, and the importance of the level of axial load increases. The significance of the level of axial loading should be examined during the development phase. C14.2.4.7.4 For the response of a structure to the design seismic shear force, building codes (e.g., UBC 97, IBC 2006 or NFPA 5000) or recommended provisions (e.g., ASCE/SEI 7-05 and FEMA 356) specify a maximum allowable drift. However, structures designed to meet that drift limit may experience greater drifts under the design basis earthquake ground motion and are likely to experience greater drifts under the risk-targeted maximum considered earthquake ground motion. In addition to the characteristics of the ground motion, actual drifts will depend on the strength of the structure, its initial elastic stiffness, and the ductility expected for the given lateral load resisting system. Specification of suitable limiting drifts for the test modules requires interpretation and allowance for uncertainties in the assumed ground motions and structural properties. In IBC 2006, the design seismic shear force applied at the base of a building is related directly to its weight and the design elastic response acceleration, and inversely to a response modification factor, R. That R factor increases with the expected ductility of the lateral force resisting system of the building. Special structural walls satisfying the requirements of Sections 21.1 and 21.9 are assigned an R value of 6 when used in a building frame system and a value of 5 when used in a bearing wall system. They are also assigned allowable story drift ratios that are dependent on the hazard to which the building is exposed. When the design seismic shear force is applied to a building, the building responds inelastically and the resultant computed drifts, (the design story drifts), must be less than a specified allowable drift. Additional guidance is given in FEMA 356 where the deformations for rectangular walls with height to length ratios greater than 2.5, and flanged wall sections with height to length ratios greater than 3.5, are to be assumed to be controlled by flexural actions. When structural walls are part of a building representing a substantial hazard to human life in the event of a failure, the allowable story drift ratio for shear controlled walls is 0.0075 and for flexure controlled walls is a function of the plastic hinge rotation at the base of the wall. For flexure controlled walls values range up to a maximum of about 0.02 for walls with confined boundary elements with low reinforcement ratios and shear stress less than 3vfc’. Figure C14.2.4.7 Example of specified test sequence. To compensate for the use of the R value, IBC Section 1617.4.6 requires that the drift determined by an elastic analysis for the code-prescribed seismic forces be multiplied by a deflection amplification factor, Cd ,to determine the design story drift and that the design story drift must be less than the allowable story drift. In building frame systems, structural walls satisfying the requirements of Section 21.9 of ACI 318 are assigned a Cd value of 5. However, research has found that design story drift ratios determined in the foregoing manner may be too low. Drift ratios of 6 times IBC-calculated values, (rather than 5), are more representative of the upper bounds to expected drift ratios. The value of 6 is also in agreement with the finding that the drift ratio of an inelastic structure is approximately the same as that of an elastic structure with the same initial period. For flexure controlled walls the value of 6/5 times the present IBC limits on calculated drift ratio, would lead to a limit on real drift ratios of up to 0.024. Duffy et al. reviewed experimental data for shear walls to define post-peak behavior and limiting drift ratios for walls with height to length ratios between 0.25 and 3.5. Seo et al. re-analyzed the data of Duffy et al. together with data from tests conducted subsequent to the analysis of Duffy et al. Duffy et al. established that for squat walls with web reinforcement satisfying ACI 318-02 requirements and height to length ratios between 0.25 and 1.1, there was a significant range of behavior for which drifts were still reliable in the post-peak response region. Typically the post-peak drift increased by 0.005 for a 20 percent degradation in capacity under cyclic loading. For greater values of degradation, drifts were less reliable. That finding has also been confirmed through tests conducted by Hidalgo et al. (2002) on squat walls with effective height to length ratios ranging between 0.35 and 1.0. Values of the drift ratio of the walls at inclined cracking and at peak capacity varied little with web reinforcement. By contrast, drifts in the post-peak range were reliable to a capacity equal to 80 percent of the peak capacity and were 0.005 greater than the drifts at peak capacity provided the walls contained horizontal and vertical web reinforcement equal to 0.25 percent. From an analysis of the available test data, and from theoretical considerations for a wall rotating flexurally about a plastic hinge at its base, Seo et al. concluded that the limiting drift at peak capacity increased almost linearly with the height to length ratio of the wall. When the additional post peak drift capacity for walls with adequate web reinforcement was added to the drift at peak capacity, the total available drift capacity in percent was given by 1.0 = 0.67 (hw / lw) + 0.5 = 3.0 where hw is the height of the wall, and lw is the length of the wall. The data from the tests of Hidalgo et al. (2002) suggest that while that formula is correct for squat walls, the lower limit on drift can be decreased to 0.8 as specified in these provisions and that the use of that formula should be limited to walls with height to length ratios equal to or greater than 0.5. For wall height to length ratios less than 0.5, the behavior is controlled principally by shear deformations (Figure C14.2.4.2.2.2c), and Equation 14.2.4-1 should not be used. The upper value of 0.030 for the drift ratio was somewhat optimistic because the data were for walls with height to length ratios equal to or less than 3.5 and subsequent tests have shown that the upper limit of 2.5, as specified in Equation 14.2.4.1, is a more realistic limit. C14.2.4.7.5 The design capacity for coupled wall systems must be developed by the drift ratio corresponding to that for the wall with the least hw/lw value. However, it is desirable that testing be continued to the drift given by Equation 14.2.4-1 for the wall with the greatest hw/lw in order to assess the reserve capacity of the coupled wall system. C14.2.4.7.6 The drift limits of Equation 14.2.4.1 are representative of the maximum that can be achieved by walls designed to ACI 318. The use of smaller drift limits is appropriate if the designer wishes to use performance measures less than the maximum permitted by ACI 318. Examples are the use of reduced shear stresses so that the likelihood of diagonal cracking of the wall is minimized or reduced compressive stresses in the boundary elements of the wall so that the risk of crushing is reduced. Nonlinear time history analyses for the response to a suite of risk-targeted maximum considered earthquakes (MCER) ground motions, rather than 1.5 times a suite of the corresponding design basis earthquake (DBE) ground motions, is required because the drifts for the response to the MCER motion can be significantly larger than 1.5 times the drifts for the response to the DBE motions. C14.2.4.7.10 In many cases, data additional to the minimum specified in Section 14.2.4.7.7 may be useful to confirm both design assumptions and satisfactory response. Such data include relative displacements, rotations, curvatures, and strains. C14.2.4.8 Test Report. The test report must be sufficiently complete and self-contained for a qualified expert to be satisfied that the tests have been designed and carried out in accordance with these criteria, and that the results satisfy the intent of these provisions. Sections 14.2.4.8.1.1 through 14.2.4.8.1.11 state the minimum evidence to be contained within the test report. The authority having jurisdiction or the registered design professional supervising the testing may require that additional test information be reported. C14.2.4.9 Test Module Acceptance Criteria. The requirements of this clause apply to each module of the test program and not to an average of the results of the program. Figure C14.2.4.9.1 illustrates the intent of this clause. Figure C14.2.4.9.1 Showing Unacceptable hysteretic behavior. C14.2.4.9.1.1 Where nominal strengths for opposite loading directions differ, as is likely for C-, L- or T- shaped walls, the criterion of Section 14.2.4.9.1.1 applies separately to each direction. C14.2.4.9.1.2 At high cyclic-drift ratios, strength degradation is inevitable. To limit the level of degradation so that drift ratio demands do not exceed anticipated levels, a maximum strength degradation of 0.20Emax is specified. Where strengths differ for opposite loading directions, this requirement applies independently to each direction. C14.2.4.9.1.3. If the relative energy dissipation ratio is less than 1/8, there may be inadequate damping for the building as a whole. Oscillations may continue for some time after an earthquake, producing low-cycle fatigue effects, and displacements may become excessive. If the stiffness becomes too small around zero drift ratio, the structure will be prone to large displacements for small lateral force changes following a major earthquake. A hysteresis loop for the third cycle between peak drift ratios of 1/10 times the limiting drift ratio given by Equation 14.2.4-1, that has the form shown in Figure C14.2.4.9.1, is acceptable. At zero drift ratio, the stiffnesses for positive and negative loading are about 11 percent of the initial stiffnesses. Those values satisfy Section 14.2.4.9.1. An unacceptable hysteresis loop form would be that shown in Figure C14.2.4.9.1 where the stiffness around zero drift ratio is unacceptably small for both positive and negative loading. COMMENTARY TO SECTION 14.4.5 C14.4.5 Modifications to Chapter 1 of ACI 530/ASCE 5/TMS 402. The seismic design factors, SDC limits, and height restrictions of these provisions are based on a combination of testing, analysis, underlying consensus standards, experience, and consistency with comparable structural systems. The testing and analysis, described in Tanner et al. (2005a and b) and Varela et al. (2005b), began in 1999 and were developed as part of an integrated research strategy. This strategy, presented at ICC-ES hearings in 2003 and affirmed in its essence using performance-based methods similar to those in the 90-percent-complete draft of FEMA P-695 (Applied Technology Council, 2008), had as its objective the development of seismic design factors consistent with at most a 10 percent probability of collapse under what was essentially equivalent to the maximum considered earthquake ground motion. That research developed factors of R and Cd equal to 3 with no restrictions on SDC or height. Additional information on that research is presented in American Society of Testing and Materials (2007), Masonry Standards Joint Committee (2005a and b and 2008a and b), The Masonry Society (2007), Tanner et al. (2005a and b), and Varela et al. (2006). Figure C14.2.4.9.1 Unacceptable hysteretic behavior. Following the initial presentation of this strategy and its associated proposals in the ICC-ES forum, it was discussed extensively with the BSSC’s Provisions Update Committee and other interested parties including the BSSC’s Code Resource Support Committee. Those discussions led to a modification of the proposal to R and Cd factors equal to 2, to SDC from A to C, and to height restrictions of 35 ft for SDC C. These values and their associated restrictions are consistent with a probability of failure much lower than 10 percent under what was essentially equivalent to the risk-targeted maximum considered earthquake ground motion (MCER). Structures of autoclaved aerated concrete (AAC) masonry are designed and constructed using U.S. consensus standards including material standards (American Society of Testing and Materials, 2007), design provisions, and mandatory construction requirements (Masonry Standards Joint Committee, 2005a and b and 2008a and b). These U.S. consensus standards are augmented by refereed documents (The Masonry Society, 2007) and the online recommendations of the Autoclaved Aerated Concrete Products Association (http://www.aacpa.org/). In the United States, AAC masonry buildings built with local approvals, under design rules consistent with the consensus standards, and with heights greater than those permitted by these provisions, have successfully resisted hurricane winds with no damage. The seismic design factors, SDC limits, and height restrictions of these provisions are consistent (or even more conservative) than those assigned to Ordinary Reinforced Masonry Shear Walls of clay or concrete masonry. ADDITIONAL REFERENCES FOR CHAPTER 14 COMMENTARY Ali, A. and J. K. Wight. 1990. Reinforced Concrete Structural Walls with Staggered Opening Configurations Under Reversed Cyclic Loading, Report UMCE 90-05. Department of Civil Engineering, University of Michigan, Ann Arbor. American Concrete Institute Innovation Task Group I and Collaborators. 2001. Acceptance Criteria for Moment Frames Based on Structural Testing, T1.1-01, and Commentary, T1.1R-01. ACI, Farmington Hills, Michigan. American Concrete Institute Innovation Task Group I and Collaborators. 2001 “Special Hybrid Moment Frames Composed of Discretely Jointed Precast and Post-Tensioned Concrete Members (ACI T1.2-XX) and Commentary (ACI T1.2R-XX),” ACI Structural Journal, 98(5):771-784. American Society of Testing and Materials. 2007. “Standard Specification for Precast Autoclaved Aerated Concrete (PAAC) Wall Construction Units,” ASTM C1386-07, Annual Book of ASTM Standards. American Society of Testing and Materials, West Conshohocken, Pennsylvania. Applied Technology Council. 2008. Quantification of Building Seismic Performance Factors: 90 Percent Complete Draft, FEMA P-695. Federal Emergency Management Agency, Washington, D.C. Banks, G., and J. Stanton. 2005. “Panel-to-Panel Connections for Hollow-Core Shear Walls Subjected to Seismic Loading,” in Proceedings of the 2005 PCI Convention, Palm Springs, California. Bora, C., M. G. Oliva, S. D. Nakaki, and R. Becker. 2005. “Development of a Precast Concrete Shear-Wall System Requiring Special Code Acceptance,” PCI Journal, 52(1):122-135. Building Seismic Safety Council. 1987. Guide to Use of NEHRP Recommended Provisions in Earthquake Resistant Design of Buildings, 1985 Edition, FEMA 140. FEMA, Washington, D.C. Building Seismic Safety Council. 2000. NEHRP Recommended Provisions for Seismic Regulations for New Buildings and Other Structures, FEMA 368 and 369. FEMA, Washington, D.C. Cheok, G. S., and H. S. Lew. 1991. “Performance of Precast Concrete Beam-to-Column Connections Subject to Cyclic Loading,” PCI Journal, 36(3):56-67. Duffy, T. A., A. Goldman, and C. R. Farrar. 1993. Shear Wall Ultimate Drift Limits, Report NUREG/CR-6104, LA-12649- MS. U.S. Nuclear Regularity Commission, Washington, D.C. Elliott, K. S., G. Davies, and W. Omar. 1992. “Experimental and Theoretical Investigation of Precast Concrete Hollow-Cored Slabs Used as Horizontal Floor Diaphragms,” The Structural Engineer, 70(10):175-187. Englekirk, R. E. 1987. “Concepts for the Development of Earthquake Resistant Ductile Frames of Precast Concrete,” PCI Journal, 32(1). Fahnestock, Larry A., James M. Ricles, and Richard Sause. 2006. “Experimental Study of a Large-Scale Buckling Restrained Using the Psudo-Dynamic Testing Method” in Proceedings of the 8th National Conference on Earthquake Engineering, San Francisco, California. Fanous, A., S. Sritharan, M. Suleiman, and A. Arulmoli, A. 2007. Minimum Spiral Reinforcement Requirements and lateral Displacement Limits for Prestressed Concrete Piles in High Seismic Regions, ISU-ERI Ames Report. Department of Civil, Construction and Environmental Engineering, Iowa State University, Ames. Federal Emergency Management Agency. 2000. “Concrete,” Chapter 6 in NEHRP Guidelines for the Seismic Rehabilitation of Buildings, FEMA 356 and 357. FEMA, Washington, D.C. French, C. W., M. Hafner, and V. Jayashanker. 1989. “Connections Between Precast Elements - Failure Within Connection Region,” ASCE Journal of Structural Engineering, 115(12):3171-3192. Hawkins, N. M., and R. E. Englekirk. 1987. “U.S.-Japan Seminar on P/C Concrete Construction in Seismic Zones,” PCI Journal, 32(2). Hidalgo, P.A., C.A. Ledezma, and R. A. Jordan. 2002. “Seismic Behavior of Squat Reinforced Concrete Shear Walls,” Earthquake Spectra, 18(2):287-308. Hutchinson, R.L., S.H. Rizkalla, M. Lau, and M. Heuvel. 1991. “Horizontal Post-Tensioned Connections for Precast Concrete Bearing Shear Walls,” PCI Journal, 36(3):64-76. International Code Council. 2006. International Building Code. ICC, Falls Church, Virginia. International Conference of Building Officials. 1997. Uniform Building Code, Vol. 2, Structural Engineering Design Provisions. ICBO, Whittier, California. Ishizuka, T., and N. M. Hawkins. 1987. Effect of Bond Deterioration on the Seismic Response of Reinforced and Partially Prestressed Concrete and Ductile Moment Resistant Frames, SM87-2. University of Washington, Department of Civil Engineering. Jayashanker, V., and C. E. French. 1988. An Interior Moment Resistant Connection Between Precast Elements Subjected to Cyclic Lateral Loads, Structural Engineering Report 87-10. University of Minnesota, Minneapolis. Kurama, Y. C. 2002. “Hybrid Post-Tensioned Precast Concrete Walls for Use in Seismic Regions,” PCI Journal, 47(5):36- 59. Masonry Standards Joint Committee (MSJC). 2005a. MSJC Code and Specification: Building Code Requirements for Masonry Structures, ACI 530-05/ASCE 5-05/TMS 402-05. The American Concrete Institute, Farmington Hills, Michigan; the American Society of Civil Engineers, Reston, Virginia; and the Masonry Society, Boulder, Colorado. Masonry Standards Joint Committee (MSJC). 2005a. MSJC Code and Specification: Specifications for Masonry Structures, ACI 530.1-05/ASCE 6-05/TMS 602-05. The American Concrete Institute, Farmington Hills, Michigan; the American Society of Civil Engineers, Reston, Virginia; and the Masonry Society, Boulder, Colorado. Masonry Standards Joint Committee (MSJC). 2008a. MSJC Code and Specification: Building Code Requirements for Masonry Structures, TMS 402-08/ACI 530-08/ASCE 5-08. The Masonry Society, Boulder, Colorado; the American Concrete Institute, Farmington Hills, Michigan; and the American Society of Civil Engineers, Reston, Virginia. Masonry Standards Joint Committee. 2008b. MSJC Code and Specification: Specifications for Masonry Structures, TMS 602-08/ACI 530.1-08/ASCE 6-08. The Masonry Society, Boulder, Colorado; the American Concrete Institute, Farmington Hills, Michigan; and the American Society of Civil Engineers, Reston, Virginia. Mast, R. F. 1992. “A Precast Concrete Frame System for Seismic Zone Four.” PCI Journal 37(1):50-64. Precast/Prestressed Concrete Institute. 2004. “Precast Prestressed Concrete Piles,” Chapter 20 in Bridge Design Manual, PCI BM-20-04. PCI, Chicago, Illinois. The Masonry Society. 2007. Masonry Designers’ Guide, 5th ed., edited by Phillip J. Samblanet. The Masonry Society, Boulder, Colorado. Nakaki, S. D., and R. E. Englekirk. 1991. “PRESSS Industry Seismic Workshops: Concept Development,” PCI Journal 36(5):54-61. Nakaki, S., J. F. Stanton, and S. Sritharan. 2001. “The PRESSS Five-Story Precast Concrete Test Building, University of California, San Diego, La Jolla, California,” PCI Journal, 46(5):20-26. National Fire Protection Association. 2006. Building Construction and Safety Code, NFPA 5000. NFPA, Quincy, Massachusetts. Neille, D. S. 1977. “Behavior of Headed Stud Connections for Precast Concrete Connections for Precast Concrete Panels Under Monotonic and Cycled Shear Loading,” thesis submitted in partial fulfillment of the requirements of Doctor of Philosophy, University of British Columbia. New Zealand Society for Earthquake Engineering. 1991. “Guidelines for the Use of Structural Precast Concrete in Buildings.” Park, R., and K. J. Thompson. 1977. “Cyclic Load Tests on Prestressed and Partially Prestressed Beam-Column Joints,” PCI Journal, 22(5):84-110. PCI Ad Hoc Committee on Precast Walls. 1997. “Design for Lateral Force Resistance with Precast Concrete Shear Walls, PCI Journal, 42(2):44-65. Pekau, O. A., and D. Hum. 1991. “Seismic Response of Friction-Jointed Precast Panel Shear Walls,” PCI Journal, 36(2):56-71. Powell, G., F. Filippou, V. Prakash, and S. Campbell. 1993. “Analytical Platform for Precast Structural Systems,” in Proceedings, ASCE Structures Congress '93. ASCE, New York. Priestley, M. J. N. 1991. “Overview of PRESSS Research Program,” PCI Journal 36(4):50-57. Priestley, M. J. N., and J. T. Tao. 1993. “Seismic Response of Precast Prestressed Concrete Frames with Partially Debonded Tendons,” PCI Journal, 38(1):58-69. Priestley, M. J. N., S. Sritharan, J. Conley, and S. Pampanin. 1999. “Preliminary Results and Conclusions from the PRESSS Five-Story Precast Concrete Test Building,” PCI Journal, 44(6):42-67. Schultz, A. E. and R.A. Magana. 1996. “Seismic Behavior of Connections in Precast Concrete Walls,” Paper SP 162-12, in Proceedings of the Mete A. Sozen Symposium, pp.273-311, ACI SP 162. American Concrete Institute, Farmington Hills, Michigan. Seo, S-Y., L-H Lee, and N. M. Hawkins. 1998. “The Limiting Drift and Energy Dissipation Ratio for Shear Walls Based on Structural Testing,” Journal of the Korean Concrete Institute,10(6):335-343. Stanton, J. F., T. R. Hicks, and N. M. Hawkins. 1991. “PRESSS Project 1.3: Connection Classification and Evaluation,” PCI Journal, 36(5):62-71. Stanton, J. F. and S. D. Nakaki. 2002. Design Guidelines for Precast Concrete Seismic Structural Systems-Unbonded Post- Tensioned Split Walls, PRESSS Report 01/03-09, UW Report SM 02-02. Department of Civil Engineering, University of Washington, Seattle. Tanner, J. E., J. L. Varela, and R. E. Klingner. 2005a. “Design and Seismic Testing of a Two-story Full-scale Autoclaved Aerated Concrete (AAC) Assemblage Specimen,” ACI Structures Journal, 102 (1):114-119. Tanner, J. E., J. L. Varela, R. E. Klingner, M. J. Brightman, and U. Cancino.2005b. “Seismic Testing of Autoclaved Aerated Concrete (AAC) Shear Walls: A Comprehensive Review,” Structures Journal, American Concrete Institute, Farmington Hills, Michigan, vol. 102, no. 3, May - June 2005, pp. 374-382. Taylor, C. P., P. E. Cote, and J. W. Wallace. “Design of Slender Reinforced Concrete Walls with Openings,” ACI Structural Journal, 95(4):420-433. Thompson, K. J., and R. Park. 1980. “Seismic Response of Partially Prestressed Concrete,” ASCE Journal of the Structural Division, 106(ST8):1755-1775. Uang, C-M., and A. Maarouf. 1993. “Seismic Displacement Amplification Factor in Uniform Building Code,” SEAONC Research Bulletin Board, BB93-3, pp. B1-B2, and “Displacement Amplification Factor for Seismic Design Provisions,” in Proceedings of the Structures Congress, ASCE, Vol.1, pp. 211-216. Irvine, California. Uriz, Patxi, and Stephen A. Mahin. 2004. “Seismic Performance Assessment of Concentrically Braced Steel Frames,” Paper 1639 in Proceedings of the 13th World Conference on Earthquake Engineering. Vancouver, B.C., Canada. Varela et al. 2006: Varela, J. L., J. E. Tanner, and R. E. Klingner. 2006. “Development of Seismic Force-Reduction and Displacement Amplification Factors for AAC Structures,” Earthquake Spectra, 22(1):267-286. Modifications to Chapter 15, Seismic Design Requirements for Nonbuilding Structures TABLE 15.4-2, SEISMIC COEFFICIENTS FOR NONBUILDING STRUCTURES NOT SIMILAR TO BUILDINGS Revise the following items as indicated (deletions in strikeout and additions underlined): Cast-in-place concrete silos, stacks, and chimneys having walls continuous to the foundation 15.6.2 3 1.75 3 NL NL NL NL NL All other reinforced masonry structures not similar to buildings 14.4.1 3 2 2.5 NL NL NL 50 50 All other nonreinforced masonry structures not similar to buildings 14.4.1 1.25 2 1.5 NL NL 50 50 50 Concrete chimneys and stacks 15.6.2 2 1.5 1.5 NL NL NL NL NL All other steel and reinforced concrete distributed mass cantilever structures not covered herein including stacks, chimneys, silos, and skirt-supported vertical vessels that are not similar to buildings 15.6.2 15.7.10 and 15.7.10.5 a and b. 3 2 2.5 NL NL NL NL NL SECTION 15.5.3, STEEL STORAGE RACKS Replace with the following: 15.5.3 Steel Storage Racks. Steel storage racks supported at or below grade shall be designed in accordance with Section 2.7 of the ANSI/RMI MH 16.1 standard and its force and displacement requirements. For storage racks supported above grade, the value of V in Section 2.7.2 of ANSI/RMI MH 16.1 shall not be taken less than the value of Fp determined in accordance with Section 13.3.1 of this standard, where Rp is taken equal to R, and ap is taken equal to 2.5. Alternatively, in addition to the requirements of Section 15.5.1, steel storage racks shall be designed in accordance with the requirements of Sections 15.5.3.1 through 15.5.3.4 [Sections 15.5.3.1 through 15.5.3.4 are unchanged.] SECTION 15.6.2, STACKS AND CHIMNEYS Replace with the following: 15.6.2 Stacks and Chimneys. Stacks and chimneys are permitted to be either lined or unlined and shall be constructed from concrete, steel, or masonry. Steel stacks, concrete stacks, steel chimneys, concrete chimneys, and liners shall be designed to resist seismic lateral forces determined from a substantiated analysis using reference documents. Interaction of the stack or chimney with the liners shall be considered. A minimum separation shall be provided between the liner and chimney equal to Cd times the calculated differential lateral drift. For concrete chimneys assigned to Seismic Design Category D, E or F, splices for vertical rebar shall be staggered such that no more than 50 percent of the bars are spliced at any elevation. Design and detailing of cross-sections in the regions of breach openings, where the loss of cross-sectional area is greater than 10 percent, shall be performed in one of the following ways: a. For vertical force, shear force, and bending moment demands along the vertical direction, design the affected crosssection using the overstrength factor of 1.5. The following detailing requirements shall be satisfied: i. The region of such overstrength shall extend above and below (except if the opening is at the base) the opening(s) by a distance equal to half of the width of the largest opening in the affected region. ii. Appropriate reinforcement development lengths shall be provided beyond the required region of overstrength. iii. The jamb regions around each opening shall be detailed using the column tie requirements in Section 7.10.5 of ACI 318. Such detailing shall extend for a jamb width of a minimum of two times the wall thickness and for a height of the opening height plus twice the wall thickness above and below the opening, but no less than the development length of the longitudinal bars. The percentage of longitudinal reinforcement in jamb regions shall meet the requirements of Section 10.9 of ACI 318 for compression members. b. Provided that the cross-sectional moment of inertia in the opening region is at least 70 percent of the same above and below it, it shall be permitted to treat the breach opening region as follows: i. All detailing requirements listed in Item a. above for the overstrength option shall be followed, in addition to the ones listed below. ii. Hoop ties in jamb regions shall be detailed as columns of intermediate moment frames using the requirements in Section 21.3.5 of ACI 318. The dimensions for jamb region shall be the same as that required in Item a. above for the overstrength option. iii. No construction joints within the opening region plus two times the wall thickness above and below the opening. iv. Ratio of outer diameter to wall thickness shall not exceed 20 within the opening region. SECTION 15.7.6, GROUND-SUPPORTED STORAGE TANKS FOR LIQUIDS Add the following exception to the end of Section 15.7.6.1, General: EXCEPTION: For Tc > 4 seconds, Sac may be determined by a site-specific study using one or more of the following methods: (a) the procedures found in Chapter 21, provided such procedures, which rely on ground-motion attenuation equations for computing response spectra, cover the natural period band containing Tc , (b) ground-motion simulation methods employing seismological models of fault rupture and wave propagation, and (c) analysis of representative strong-motion accelerogram data with reliable long-period content extending to periods greater than Tc . However, in no case shall the value of Sac be taken as less than the minimum of: 1. The value determined in accordance with Equation 15.7-11 using 50 percent of the mapped value of TL from Figure 22-7 or 2. 0.8 times the value determined in accordance with Equation 15.7-11 using the mapped value of TL from Figure 22-7. In determining the value of Sac, the value of TL shall not be less than 4 seconds. Commentary to Chapter 15 Modifications COMMENTARY TO SECTION 15.5.3 C15.5.3 Steel Storage Racks. The two approaches to the design of steel storage racks set forth by the standard are intended to produce comparable results. The specific revisions to the RMI specification cited in earlier editions of the Provisions and the detailed requirements of the new ANSI/RMI standard reflect the recommendations of FEMA 460, Seismic Considerations for Steel Storage Racks Located in Areas Accessible to the Public. COMMENTARY TO SECTION 15.6.2 C15.6.2 Stacks and Chimneys. The design of stacks and chimneys to resist natural hazards generally is governed by wind design considerations. The exceptions to this general rule involve locations with high seismicity, stacks and chimneys with large elevated masses, and stacks and chimneys with unusual geometries. It is prudent to evaluate the effect of seismic loads in all but those areas with the lowest seismicity. Although not specifically required, it is recommended that the special seismic details required elsewhere in the standard be considered for application to stacks and chimneys. Concrete chimneys have low ductility, and their seismic behavior is especially critical in the opening regions due to inherent reduction in strength and loss of confinement for vertical reinforcement in the jamb regions around the openings. Spectacular earthquake-induced chimney failures have occurred in recent history (in Turkey in 1999) and have been attributed to strength/detailing problems (Kilic and Sozen, 2003). Therefore, the R value of 3 traditionally used in ASCE/SEI 7-05 for concrete stacks and chimneys is reduced to 2 and detailing requirements for breach openings are added in the 2009 NEHRP Recommended Seismic Provisions. Guyed steel stacks and chimneys are generally lightweight. As a result, the design loads due to natural hazards generally are governed by wind. On occasion, large flares or other elevated masses located near the top may require in-depth seismic analysis. Although it does not specifically address seismic loading, Chapter 6 of Troitsky (1982) provides a methodology appropriate for resolution of the seismic forces defined in the standard. COMMENTARY TO SECTION 15.7.6.1 C15.7.6.1 General. The response of ground storage tanks to earthquakes is well documented by Housner, Mitchell and Wozniak, Veletsos, and others. Unlike building structures, the structural response of these tanks is influenced strongly by the fluid-structure interaction. Fluid-structure interaction forces are categorized as sloshing (convective) and rigid (impulsive) forces. The proportion of these forces depends on the geometry (height-to-diameter ratio) of the tank. API 650, API 620, AWWA D100, AWWA D110, AWWA D115, and ACI 350.3 provide the data necessary to determine the relative masses and moments for each of these contributions. The standard requires that these structures be designed in accordance with the prevailing reference documents, except that the height of the sloshing wave, ds , must be calculated using Equations 15.7-13. Note that API 650 and AWWA D100 include this requirement in their latest editions. Equations 15.7-10 and 15.7-11 provide the spectral acceleration of the sloshing liquid for the constant-velocity and constantdisplacement regions of the response spectrum, respectively. The 1.5 factor in these equations is an adjustment for 0.5 percent damping. An exception in the use of Equation 15.7-11 was added for the 2009 NEHRP Recommended Seismic Provisions. Actual site-specific studies carried out since the introduction of the TL requirements of ASCE/SEI 7-05 indicate that the mapped values of TL are extremely conservative. Because a revision of the TL maps is a time-consuming task that would not be possible during the 2009 Provisions update cycle, an exception was added to allow the use of site-specific values that are less than the mapped values with a floor of 4 seconds or one-half the mapped value of TL. The exception was added under Section 15.7.6 because TL is a tank issue. Discussion of the site-specific procedures can be found in the Part 2 Commentary for Chapter 22. ADDITIONAL REFERENCE FOR CHAPTER 15 COMMENTARY Kilic, S., and M. Sozen. 2003. “Evaluation of Effect of August 17, 1999, Marmara Earthquake on Two Tall Reinforced Concrete Chimneys,” ACI Structural Journal, 100(3). Page intentionally left blank. Modification to Chapter 16, Seismic Response History Procedures SECTION 16.1.3.2, THREE-DIMENSIONAL ANALYSIS Replace with the following: 16.1.3.2 Three-Dimensional Analysis. Where three-dimensional analyses are performed, ground motions shall consist of pairs of appropriate horizontal ground motion acceleration components that shall be selected and scaled from individual recorded events. Appropriate ground motions shall be selected from events having magnitudes, fault distances, and source mechanisms that are consistent with those that control the risk-targeted maximum considered earthquake (MCER). Where the required number of recorded ground motion pairs is not available, appropriate simulated ground motion pairs are permitted to be used to make up the total number required. For each pair of horizontal ground motion components, a square root of the sum of squares (SRSS) spectrum shall be constructed by taking the SRSS of the 5-percent-damped response spectra for the scaled components (for direct scaling, an identical scale factor is applied to both components of a pair). Each pair of motions shall be scaled such that for each period between 0.2T and 1.5T, the average of the SRSS spectra from all horizontal component pairs does not fall below the corresponding ordinate of the MCER response spectrum determined in accordance with Section 11.4.5 or 11.4.7. At sites within 5 km of an active fault that controls the hazard, each pair of components shall be rotated to the faultnormal and fault-parallel direction of the causative fault and shall be scaled so that the average of the fault-normal components is not less than the MCER response spectrum for each period between 0.2T and 1.5T. Commentary to Chapter 16 Modification COMMENTARY TO SECTION 16.1.3.2 C16.1.3.2 Three-dimensional Analyses. One key change to the ground motion design requirements developed by the BSSC’s Seismic Design Procedure Review Group (SDPRG) for the 2009 NEHRP Recommended Seismic Provisions is the use of maximum direction ground motions. In addition to changing the design values defined in Chapter 11 and used throughout the Provisions, implementing maximum direction ground motions affects the previous ground motion scaling rules specified in Section 16.1.3.2. Studies (Maffei and Hashemi, 2008) of 50 ground motions of M6.5-M7.9 earthquakes for both far-field and near-field records and for periods in the range of 0.1 to 3.0 seconds indicate that the maximum direction of ground motion is slightly less than the SRSS of the two components with the SRSS spectrum tending to be approximately 1.16 times the maximum direction spectrum. For each of the 50 ground motions, the maximum response of a single-degree-of-freedom (SDOF) oscillator (assuming 5- percent damping) was determined for ground motion orientations from 0 to 90 degrees (in one-degree increments) and was compared to the associated SRSS of maximum response. The ratios of the SRSS of maximum response and the maximum amplitude of the response for varying parameters are given in Tables C16.1.3.2-1 through C16.1.3.2-3. Table C16.1.3.2-1 Ratio of SRSS of Maximum Response to Maximum Amplitude as a Function of SDOF Period SDoF Period Number of Data Points Ratio-Mean Ratio –Standard Deviation 0.1 sec 50 1.19 0.077 0.3 sec 50 1.16 0.068 1.0 sec 50 1.14 0.067 3.0 sec 50 1.13 0.077 Average 200 1.16 0.076 Table C16.1.3.2-2 Ratio of SRSS of Maximum Response to Maximum Amplitude as a Function of Ground Motion Records Ground Motion Number of Data Points Ratio-Mean Ratio -Standard Deviation Far-Field 88 1.16 0.067 Near-Field 112 1.15 0.078 Average 200 1.16 0.076 Table C16.1.3.2-3 Ratio of SRSS of Maximum Response to Maximum Amplitude as a Function of Site Class Site Class Number of Data Points Ratio-Mean Ratio -Standard Deviation B 8 1.15 0.066 C 84 1.15 0.072 D 108 1.16 0.073 Average 200 1.16 0.076 The modified scaling requirements simplify phrasing of existing language by replacing 10 percent less than 1.16 times the MCER response spectrum with the MCER response spectrum, itself, resulting in an effective “1.0” multiplier. This effective multiplier comes from (0.9)(1.16) ˜ 1.0. However, for sites within approximately 5 km of an active fault that controls the ground-motion hazard, the near field strongmotion database indicates that the fault-normal (FN) direction is (or is close to) the direction of maximum ground motion for periods around 1.0 second and greater (Huang et al., 2008; Watson-Lamprey and Boore, 2007). In this case, the two horizontal components of a selected record are to be transformed so that one component is the motion in the FN direction and the other component is the motion in the fault-parallel (FP) direction. Scaling so that the average FN component response spectrum is at the level of the MCER response spectrum ensures that the FN components will not be underestimated, which would happen if the SRSS rule was applied at short distances. The same scale factor selected for the FN component of a given record is used for the FP component also. ADDITIONAL REFERENCES FOR CHAPTER 16 COMMENTARY Huang, Y. N., A. Whittaker, and N. Luco. 2007. “NGA Relationships, USGS Seismic Hazard Maps, Near-Fault Ground Motions and Site Effects: BSSC Project 07 Final Draft Report. BSSC, Washington, D.C. Maffei, J., and A. Hashemi. 2008. Personal Communication. Watson-Lamprey, J. A., and D. M. Boore. 2007. “Beyond SaGMRotI: Conversion to SaArb, SaSN, and SaMaxRot,” Bulletin of the Seismology Society of America, 97:1511-1524. Modifications to Chapter 18, Seismic Design Requirements for Structures with Damping Systems SECTION 18.3.1, NONLINEAR RESPONSE HISTORY PROCEDURE Replace with the following: 18.3.1 Nonlinear Response History Procedure. A nonlinear response history (time history) analysis shall utilize a mathematical model of the structure and the damping system as provided in Chapter 16 and this section. The model shall directly account for the nonlinear hysteretic behavior of elements of the structure and the damping devices to determine its response, through methods of numerical integration, to suites of ground motions compatible with the design response spectrum for the site. The analysis shall be performed in accordance with Chapter 16 together with the requirements of this section. Inherent damping of the structure shall not be taken greater than 5 percent of critical unless test data consistent with levels of deformation at or just below the effective yield displacement of the seismic-force-resisting system support higher values. If the calculated force in an element of the seismic force-resisting system does not exceed 1.5 times its nominal strength, that element is permitted to be modeled as linear. 18.3.1.1 Damping Device Modeling. Mathematical models of displacement-dependent damping devices shall include the hysteretic behavior of the devices consistent with test data and accounting for all significant changes in strength, stiffness, and hysteretic loop shape. Mathematical models of velocity-dependent damping devices shall include the velocity coefficient consistent with test data. If this coefficient changes with time and/or temperature, such behavior shall be modeled explicitly. The elements of damping devices connecting damper units to the structure shall be included in the model. Exception: If the properties of the damping devices are expected to change during the duration of the response history analysis, the dynamic response is permitted to be enveloped by the upper and lower limits of device properties. All these limit cases for variable device properties must satisfy the same conditions as if the time dependent behavior of the devices were explicitly modeled. 18.3.1.2 Response Parameters. For each ground motion analyzed, individual response parameters consisting of the maximum value of the individual member forces, member inelastic deformations and story drifts at each story shall be determined. Moreover, for each ground motion used for response history analysis, individual response parameters consisting of the maximum value of the discrete damping device forces, displacements, and velocities, in the case of velocity-dependent devices, shall be determined. If at least seven ground motions are used for response history analysis, the design values of the damping device forces, displacements, and velocities are permitted to be taken as the average of the values determined by the analyses. If fewer than seven ground motions are used for response history analysis, the design damping device forces, displacements and velocities shall be taken as the maximum value determined by the analyses. A minimum of three ground motions shall be used. SECTION 18.3.2, NONLINEAR STATIC PROCEDURE Replace with the following: 18.3.2 Nonlinear Static Procedure. Nonlinear static procedures may be used to construct the lateral forcedisplacement curve of the seismic-force-resisting system in lieu of the elastoplastic curve assumed in the response spectrum procedure and in the equivalent lateral force procedure. When nonlinear static procedures is used, the nonlinear modeling described Chapter 16 shall be used. The resulting force-displacement curve shall be used in lieu of the assumed effective yield displacement, DY, of Equation 18.6-10 to calculate the effective ductility demand under the design earthquake ground motion, µD, and under the risk-targeted maximum considered earthquake ground motion, µM, in Equations 18.6-8 and 18.6-9, respectively. The value of (R/Cd) shall be taken as 1.0 in Equations 18.4-4, 18.4-5, 18.4- 8, and 18.4-9 for the response spectrum procedure, and in Equations 18.5-6, 18.5-7 and 18.5-15 for the equivalent lateral force procedure. Page intentionally left blank. FIGURE 19.2-1, FOUNDATION DAMPING FACTOR Chapter 19, Soil Structure Interaction for Seismic Design TABLE 19.2-1, VALUES OF G / G0 AND Vs / Vso Replace with the following: Site Class Value of vs / vs0 Value of G / G0 SDS/2.5 = 0.1 0.4 = 0.8 = 0.1 0.4 = 0.8 A 1.00 1.00 1.00 1.00 1.00 1.00 B 1.00 0.97 0.95 1.00 0.95 0.90 C 0.97 0.87 0.77 0.95 0.75 0.60 D 0.95 0.71 0.32 0.90 0.50 0.10 E 0.77 0.22 * 0.60 0.05 * F * * * * * * Note: Use straight line interpolation for intermediate values of SDS/2.5. * Should be evaluated from site-specific analysis. FIGURE 19.2-1, FOUNDATION DAMPING FACTOR Replace with the following: Page intentionally left blank. Modification to Chapter 21, Site-Specific Ground Motion Procedures for Seismic Design SECTION 21.2, GROUND MOTION HAZARD ANALYSIS Replace Sections 21.2.1 through 21.2.3 with the following: 21.2.1 Probabilistic Ground Motions. The probabilistic spectral response acceleration shall be taken as the spectral response acceleration in the maximum direction of ground motions represented by a 5 percent damped acceleration response spectrum that is expected to achieve a 1 percent probability of collapse within a 50-year period. For the purpose of this provision, ordinates of the probabilistic ground-motion response spectrum shall be determined by either Method 1 of Section 21.2.1.1 or Method 2 of Section 21.2.1.2. 21.2.1.1 Method 1. Ordinates of the probabilistic ground-motion response spectrum shall be determined as the product of the risk coefficient at each spectral response period, CR, and the spectral response acceleration represented by a 5 percent damped acceleration response spectrum having a 2 percent probability of exceedance within a 50-year period. The value of the risk coefficient, CR, shall be determined using values of CRS and CR1 from Figures 22-3 and 22-4, respectively. At spectral response periods less than or equal to 0.2 second, CR shall be taken as equal to CRS. At spectral response periods greater than or equal to 1.0 second, CR shall be taken as equal to CR1. At response spectral periods greater than 0.2 second and less than 1.0 second, CR shall be based on linear interpolation of CRS and CR1. 21.2.1.2 Method 2. Ordinates of the probabilistic ground-motion response spectrum shall be determined at each spectral response period from the iterative integration of a site-specific hazard curve with a lognormal probability density function representing the collapse fragility (i.e., probability of collapse as a function of spectral response acceleration). At each period, the ordinate of the probabilistic ground-motion response spectrum shall achieve a 1 percent probability of collapse within a 50-year period for a collapse fragility having (i) a 10 percent probability of collapse at said ordinate of the probabilistic ground-motion response spectrum and (ii) a logarithmic standard deviation value of 0.8. 21.2.2 Deterministic Ground Motions. The deterministic spectral response acceleration at each period shall be calculated as the largest 84th percentile 5 percent damped spectral response acceleration in the direction of maximum horizontal response computed at that period for characteristic earthquakes on all known active faults within the region. For the purposes of this standard, the ordinates of the deterministic ground motions response spectrum shall not be taken as lower than the corresponding ordinates of the response spectrum determined in accordance with Figure 21.2-1, where Fa and Fv are determined using Tables 11.4-1 and 11.4-2, respectively, with the value of Ss taken as 1.5 and the value of S1 taken as 0.6. 21.2.3 Site-Specific MCER. The site-specific MCER spectral response acceleration at any period, SaM, shall be taken as the lesser of the spectral response accelerations from the probabilistic ground motions of Section 21.2.1 and the deterministic ground motions of Section 21.2.2. Commentary to the Chapter 21 Modification C21.2 GROUND MOTION HAZARD ANALYSIS As explained in the commentary to Chapter 11, the risk-targeted maximum considered earthquake ground motions (MCER) in the 2009 NEHRP Recommended Seismic Provisions are based on the 2008 USGS seismic hazard maps and also incorporate three technical changes to ASCE/SEI 7-05: 1. Use of risk-targeted ground motions, 2. Use of maximum direction ground motions, and 3. Use of near-source 84th percentile ground motions. Reasons for use of maximum direction ground motions are explained first in the commentary below, because they apply to both the probabilistic and deterministic ground motions discussed subsequently. Use of risk-targeted and near-source 84th percentile ground motions are discussed in the probabilistic and deterministic ground motions sections below, respectively. The requirements in the previous editions of the Provisions and ASCE/SEI 7 do not define the direction of ground motions used for design. The procedure used to develop the statistical estimate of ground motion results in the geometric mean (geomean) of two orthogonal components of motion at a site. Many engineers find the maximum direction to be a more meaningful parameter for structural design. The basic concept is that a structure is designed to resist the ground motion at its site; the prediction of ground motion is inherently statistical, and the basis for the statistical estimate of the ground motion is rooted in the probability that a structure will actually fail. In general, structures will not have the same resistance in all directions; however, for those structures in which seismic resistance is a significant economic factor, there is a tendency to design to the level required by building codes, with the result that the resistance of the structure is relatively insensitive to the direction of the motion. When one considers such structures subjected to two simultaneous components of ground motion, these structures characteristically fail in the direction of the stronger of the two components. Failure rates of simple buildings in one recent study (low-rise wood buildings in Applied Technology Council, 2008) show this effect: the overall failure rate for three-dimensional analyses was higher than those for two-dimensional analyses for the same set of structures analyzed for the same 22 pairs of ground motions. The specification of maximum direction ground motions reduces the probability of structural failure based upon equivalent static two-dimensional design compared to the use of the geomean based demand, but this reduction has not been quantified at this time. For consistency, revisions have been made to both probabilistic and deterministic ground motion criteria to reflect required use of maximum direction ground motions. The USGS updates of the uniform-hazard and deterministic ground motion spectral value maps have used the new next generation attenuation (NGA) relations for sites in the western United States (WUS). The new NGA relationships output an average horizontal spectral demand and the dispersion in that demand, where this average is the rotated geomean denoted as GMRotI50 (Boore et al., 2006). GM denotes the geometric mean of two horizontal components, Rot denotes that rotations over all non-redundant angles are considered, I denotes that period-independent rotations are used, and 50 identifies the prediction of median values. The geometric mean of two horizontal components of ground motions is calculated as the square root of the product of the two horizontal response spectral accelerations at each period of interest. As demonstrated by Boore et al. (2006), GMRotI50 is numerically very similar to (i.e., within 3 percent of) the geometric mean of two asrecorded components that was typically the output of older attenuation relationships. A recent study (Huang et al., 2008a) found that near-source ground motion spectral response accelerations of the new NGA relations are somewhat less than those in the maximum direction of response. This study (2008a, 2008b) focused on large magnitude earthquakes, with moment magnitudes greater than 6.5 and site-to-source distances less than 15 km. For this family of earthquake records, ground motions in the maximum direction of response are about 110 percent of 5 percent damped, short-period spectral response acceleration, and about 130 percent of 5 percent damped, 1-second spectral response acceleration calculated using the new NGA relations (GMRotI50). Table C21.2-1 presents summary results to enable calculation of median and 84th percentile ratios of maximum to geomean spectral demands across the period range of 0 to 4.0 seconds; values of the ratio are assumed to remain constant for periods greater than 4.0 seconds. Values are rounded to the nearest 0.1, which is the appropriate degree of precision. The ratio of 84th percentile (Column 3) to median (Column 2) demands is approximately 1.8 to 1.9. Linear interpolation should be used to establish values of the ratios for periods not listed. Other regions (e.g., the central and eastern United States) are expected to have similar ratios of maximum direction ground motions to geomean ground motions although the limited number of strong-motion records from the central and eastern United States precludes rigorous evaluation such as that performed by the NGA study (Huang et al., 2008). However, studies by Beyer and Bommer (2006) using a set of 949 earthquake records with much wider ranges of moment magnitude (4.2 to 7.9) and hypocentral distance (5 to 200 km) indicated similar ratios of maximum to geomean response to those of the Huang et al. study on large magnitude, near-fault ground motions. The Beyer and Bommer data set included records from 20+ European earthquakes. Table C21.2-1 Median and 84th Percentile of the Ratio of Maximum Spectral Demand to Geomean Demand Period (second) Median 84th Percentile Period (second) Median 84th Percentile 0.0 1.1 2.0 0.5 1.2 2.1 0.1 1.1 2.0 1.0 1.3 2.3 0.2 1.1 2.0 2.0 1.3 2.5 0.3 1.1 2.0 4.0+ 1.4 2.7 For consistency of ground motion scaling (against either geomean or maximum direction spectra) in three-dimensional response history analysis of structures, the 2009 Provisions has adopted changes related to Section 16.1.3.2 of ASCE/SEI 7- 05 such that it enables the scaling of pairs of horizontal ground motion records matching maximum direction spectra (MCER or design spectra of maximum direction of response) to be equivalent to that matching the corresponding geomean spectra. Additional explanation of these changes is provided in Section C16.1.3.2. C21.2.1 Probabilistic Ground Motions. The definition and basis of probabilistic ground motions in these new Provisions has changed from that in ASCE/SEI 7-05, from a 2 percent in 50-year hazard level to a 1 percent in 50-year collapse risk target. This change is intended to improve seismic design by achieving a more uniform level of collapse prevention. The change affects the calculation and values of probabilistic ground motions, but not their use in the design process (i.e., 5 percent damped spectral response accelerations are still used). The technical basis of the change can be found in “Risk- Targeted versus Current Seismic Design Maps for the Conterminous United States” (Luco et al., 2007). A summary of the technical basis is provided below. In the 1997, 2000 and 2003 editions of the NEHRP Recommended Provisions, the probabilistic MCE ground motions are defined as those that have a 2 percent probability of being exceeded in 50 years. In other words, the probabilistic MCE ground motions are of uniform hazard, both geographically and across structural vibration periods. It has long been recognized, however, that “it really is the probability of structural failure with resultant casualties that is of concern, and the geographical distribution of that probability is not necessarily the same as the distribution of the probability of exceeding some ground motion” (p. 296 of ATC 3-06, 1978). The primary reason that the two probabilities are not the same is that there are geographic differences in the shape of the ground motion versus annual frequency of exceedance hazard curves from which uniform-hazard ground motions are read. The commentary of earlier editions of the Provisions (post-1997) reports that “because of these differences, questions were raised concerning whether definition of the ground motion based on a constant probability for the entire United States would result in similar levels of seismic safety for all structures” (p. 319 of the 2003 NEHRP Recommended Provisions Commentary). The change to risk-targeted ground motions uses the different shapes of hazard curves to adjust the uniformhazard (2-percent-in-50-years) ground motions such that they are expected to result in a uniform annual frequency of collapse, or risk level, when used in design. The adjustment factors, or risk coefficients, are akin to the ASCE/SEI 43-05 site-specific design factor, which is a function of an approximate slope of the ground motion hazard curve. The adjustments to the uniform-hazard ground motions are computed by making use of the so-called risk integral (e.g., McGuire, 2004). The risk integral calculates an annual frequency of collapse by coupling the ground motion hazard curve at a location with the expected performance of a structure designed for that location. More precisely, the hazard curves are coupled with the conditional probability of collapse as a function of the ground motion level. Earlier editions of the Provisions express the expectation that “if a structure experiences a level of ground motion 1.5 times the design level [i.e., the MCE ground motion], the structure should have a low likelihood of collapse” (p. 320 of the 2003 NEHRP Provisions Commentary). This “low likelihood of collapse” has been estimated as 10 percent (Applied Technology Council, 2009) using state-of-the-art incremental dynamic analysis (e.g., Vamvatsikos and Cornell, 2002) of structures designed in accordance with this edition of the NEHRP Recommended Seismic Provisions (2009). For the likelihood of collapse under other (than the MCE) ground motion levels, a so-called ß-value of 0.8 has been used for the 2009 Provisions, based on both the findings of the Applied Technology Council (2009) and other past research. Other ß-values ranging from 0.5 to 1.0 have been considered, with little effect on the resulting risk coefficients. The ground motion hazard curves used in the risk integral are from the USGS. Using more subjective estimates of the conditional probability of collapse as a function of the ground motion level, and early (1976) hazard curves for only four locations, the authors of the resource document on which the Provisions are based (Applied Technology Council, 1978) used the risk integral to estimate the annual frequency of collapse of buildings designed for uniform-hazard ground motions (see ATC 3-06, p. 310-311). They found that “the probabilities of failure [i.e., risk levels] were roughly the same for each of the four buildings.” In contrast, using contemporary hazard curves and building performance expectations, Luco et al. (2007) have found that the risk levels are systematically lower in the central and eastern United States (CEUS) than in the WUS due to well-documented differences in the shapes of ground motion hazard curves (e.g., Leyendecker et al., 2000). To result in uniform risk levels, adjustments to the uniform-hazard ground motions are needed. The risk level targeted in these Provisions (2009) corresponds (approximately) to 1 percent probability of collapse in 50 years. This target is based on the average of the annual frequencies of collapse across the WUS that are expected to result from (as calculated via the risk integral) design for the probabilistic MCE ground motions in the 2003 NEHRP Recommended Provisions. Consequently, the requisite risk coefficients are generally within 15 percent of unity in the WUS (except in the coastal region of Oregon, where they are slightly smaller). In the CEUS, the risk coefficients are generally smaller, again due to the well-documented differences in shapes of ground motion hazard curves there relative to the WUS. In the New Madrid seismic zone and near Charleston, South Carolina, in particular, the adjustments to the uniform-hazard ground motions are as small as a factor of 0.7. Compared to the underlying uniform-hazard ground motions, the risk coefficients are generally less sensitive to refinements of the ground motion hazard curves (e.g., USGS updates or site-specific analyses), since they depend on the shape but not amplitude of the hazard curves. They vary with the structural vibration period and site class, but not dramatically. The change to risk-targeted probabilistic ground motions complements improvements to the basis for response modification factors (R factors) reflected in FEMA P-695 (Applied Technology Council, 2009) and provides a more rational basis for seismic design methods. As alluded to above, similar risk-based procedures are already being used for design and evaluation of nuclear facilities, as well as offshore structures. C21.2.2 Deterministic Ground Motions. Deterministic ground motions should account for uncertainties associated with near-fault ground motions, particularly at longer periods, and necessitate a more statistically appropriate estimate of 5 percent damped spectral response accelerations than those based on the 150 percent of the median ground motions used in ASCE/SEI 7-05. The use of 84th percentile ground motions in these Provisions (2009) effectively requires increasing median ground motions by 180 percent. The technical basis of this change can be found in Huang et al. (2008a and 2008b). The authors found that 150 percent of the median spectral response accelerations of the new NGA relations (average of the three relations) to be significantly less than 84th percentile ground motions in the maximum direction of response. Near active sources (in the WUS), 84th percentile ground motion in the maximum direction of response is about 200 percent (1.8 x 110 percent) of 5 percent damped, short-period spectral response acceleration, and about 230 percent (1.8 x 130 percent) of 5 percent damped, 1-second spectral response acceleration of the new NGA relations for GMRotI50 (average value of the three NGA relations). Table C21.2-2 summarizes ratios of 84th percentile maximum direction to median geomean-direction response for periods from 0 to 4.0 seconds. Ratios for periods greater than 4.0 seconds are assumed to be the same as the ratio for 4.0 seconds. Table C21.2-2 Ratios of 84th Percentile to Median Spectral Demands for NGA Relationships Period (seconds) 0.2 0.5 1.0 2.0 3.0 4.0 Equation ß B-A 0.60 0.61 0.65 0.70 0.70 0.70 C-B 0.59 0.59 0.62 0.64 0.65 0.65 C-Y 0.61 0.63 0.63 0.67 0.67 0.70 Equation y84 / y50 1.82 1.84 1.89 1.95 1.96 1.98 ADDITIONAL REFERENCES FOR CHAPTER 21 COMMENTARY Abrahamson, N., and W. J. Silva. 1997. "Empirical Response Spectral Attenuation Relations for Shallow Crustal Earthquakes." Seismological Research Letters, 68(1):94-127. American Society of Civil Engineers. 2006. Minimum Design Loads for Buildings and Other Structures, ASCE/SEI 7-05. ASCE, Reston, Virginia. American Society of Civil Engineers. 2005. Seismic Design Criteria for Structures, Systems, and Components in Nuclear Facilities, ASCE 43-05. ASCE, Reston, Virginia. Applied Technology Council. 1978. Tentative Provisions for the Development of Seismic Regulations for Buildings, ATC 3- 06. ATC, Palo Alto, California. Applied Technology Council. 2009. Quantification of Building Seismic Performance Factors, FEMA P-695. Federal Emergency Management Agency, Washington, D.C. Beyer, K., and J. J. Bommer. 2006. Relationships between Median Values and between Aleatory Variabilities for Different Definitions of the Horizontal Component of Motion, Bulletin of the Seismological Society of America, 96(4A):1512-1522. Boore, D. M., and T. E. Fumal. 1997. "Equations for Estimating Horizontal Response Spectra and Peak Acceleration from Western North American Earthquakes: A Summary of Recent Work," Seismological Research Letters, 68(1):128-153. Boore, D. M., J. Watson-Lamprey, and N. A. Abrahamson. 2006. "Orientation-Independent Measures of Ground Motion," Bulletin of the Seismological Society of America, 96(4A):1502-1511. Boore, D. M., and G. M. Atkinson. 2007. Boore-Atkinson NGA Ground Motion Relations for the Geometric Mean Horizontal Component of Peak And Spectral Ground Motion Parameters, PEER 2007/01. Pacific Earthquake Engineering Research Center, Berkeley, California. Campbell, K. W., and Y. Bozorgnia. 2003. "Updated Near-Source Ground Motion (Attenuation) Relations for the Horizontal and Vertical Components of Peak Ground Acceleration and Acceleration Response Spectra." Bulletin of the Seismological Society of America, 93:314-331. Campbell, K. W., and Y. Bozorgnia. 2007. Campbell-Bozorgnia NGA Ground Motion Relations for the Geometric Mean Horizontal Component of Peak and Spectral Ground Motion Parameters," PEER 2007/02. Pacific Earthquake Engineering Research Center, Berkeley, California. Chiou, B. S.-J., and R. R. Youngs. 2006. Chiou and Youngs PEER-NGA Empirical Ground Motion Model for the Average Horizontal Component of Peak Acceleration and Pseudo-Spectral Acceleration for Spectral Periods of 0.01 to 10 Seconds. Pacific Earthquake Engineering Research Center, Berkeley, California, http://peer.berkeley.edu/products/CYProgram/ Chiou_Youngs_NGA_2006.pdf. Huang, Y.-N, A. S. Whittaker, and N. Luco. 2008a. NGA Relationships, USGS Seismic Hazard Maps, Near-Fault Ground Motions and Site Effects, USGS, Golden, Colorado. Huang, Y.-N, A. S. Whittaker, and N. Luco. 2008b. “Maximum Spectral Demands in the Near-Fault Region,” Earthquake Spectra, 24(1):319-341. Leyendecker, E.V., R. J. Hunt, A. D. Frankel, and K. S. Rukstales. 2000. “Development of Maximum Considered Earthquake Ground Motion Maps,” Earthquake Spectra, 16(1):21-40. Luco, N., B. R. Ellingwood, R. O. Hamburger, J. D. Hooper, J. K. Kimball, and C. A. Kircher. 2007. “Risk-Targeted versus Current Seismic Design Maps for the Conterminous United States,” in Proceedings of the SEAOC 76th Annual Convention. Structural Engineers Association of California, Sacramento, California. McGuire, R. K. 2004. Seismic Hazard and Risk Analysis,” EERI Monograph MNO-10. Earthquake Engineering Research Institute, Oakland, California. Sadigh, K., C. Y. Chang, J. A. Egan, F. Makdisi, and R. R. Youngs. 1997. "Attenuation Relationships for Shallow Crustal Earthquakes Based on California Strong Motion Data." Seismological Research Letters, 68(1):180-189. Somerville, P. G., N. F. Smith, R. W. Graves, and N. A. Abrahamson. 1997. "Modification of Empirical Strong Ground Motion Attenuation Relations To Include the Amplitude And Duration Effects of Rupture Directivity," Seismological Research Letters, 68(1):199-222. Vamvatsikos, D., and C. A. Cornell. 2002. “Incremental Dynamic Analysis,” in Earthquake Engineering and Structural Dynamics, 31(3):491-514. Page intentionally left blank. Modification to Chapter 22, Seismic Ground Motion and Long-period Transition Maps Replace existing Chapter 22 with the following: Chapter 22 SEISMIC GROUND MOTION, LONG-PERIOD TRANSITION, RISK COEFFICIENT, AND MCE GEOMEAN PGA MAPS Contained in this chapter are Figures 22-1 through 22-7, which provide the mapped uniform-hazard ground motion parameters (SSUH and S1UH), the mapped risk coefficients (CRS and CR1), the mapped deterministic ground motion parameters (SSD and S1D), and the mapped long-period transition period (TL), for use in applying the seismic provisions of ASCE/SEI 7. Also contained in this chapter are Figures 22-8 through 22-11, which provide the mapped maximum considered earthquake geometric mean peak ground accelerations. These maps were prepared by the United States Geological Survey (USGS) and have been updated for the 2009 edition of the NEHRP Recommended Seismic Provisions for New Buildings and Other Structures. Maps for Guam and Tutuila (American Samoa) are not included because uniform-hazard ground motion parameters, deterministic ground motion parameters, and risk coefficients have not yet been developed for those islands. Therefore, like in the 2005 edition of ASCE/SEI 7, the parameters SS and S1 defined in Section 11.4.3 shall be, respectively, 1.5 and 0.6 for Guam and 1.0 and 0.4 for Tutuila. The mapped maximum considered earthquake geometric mean peak ground accelerations shall be 0.6 for Guam and 0.4 for Tutuila. The following is a list of figures contained in this chapter: Figure 22-1 Uniform-hazard (2% in 50-year) ground motions of 0.2-second spectral response acceleration (5% of critical damping), Site Class B. Figure 22-2 Uniform-hazard (2% in 50-year) ground motions of 1-second spectral response acceleration (5% of critical damping), Site Class B. Figure 22-3 Risk coefficient at 0.2-second spectral response period. Figure 22-4 Risk coefficient at 1-second spectral response period. Figure 22-5 Deterministic ground motions of 0.2-second spectral response acceleration (5% of critical damping), Site Class B. Figure 22-6 Deterministic ground motions of 1-second spectral response acceleration (5% of critical damping), Site Class B. Figure 22-7 Long-period transition period, TL (seconds). Figure 22-8 MCE geometric mean PGA, %g, Site Class B for the coterminous United States. Figure 22-9 MCE geometric mean PGA, %g, Site Class B for Alaska. Figure 22-10 MCE geometric mean PGA, %g, Site Class B for Hawaii. Figure 22-11 MCE geometric mean PGA, %g, Site Class B for Puerto Rico and the United States Virgin Islands. Figure22-1 Map of the US Figure22-1 (cont) Map of the US Figure 22-2 Map of the US Figure 22-2 (cont) Map of the US Figure 22-3 Map of the US Figure 22-3 (cont) Map of the US Figure 22-4 Map of the US Figure 22-4 (cont) Map of the US Figure 22-5 Map of the US Figure 22-5 (cont) Map of the US Figure 22-6 Map of the US Figure 22-6 (cont) Map of the US Figure 22-7 Map of the US Figure 22-7 (cont) Map of the US Figure 22-8 Map of the US Figure 22-8 (cont) Map of the US Figures 22-9 and 10 Map of the US Figures 22-11 showing a US Map Commentary to New Chapter 22 Chapter 22 Commentary SEISMIC GROUND MOTION, LONG-PERIOD TRANSITION, RISK COEFFICIENT, AND MCE GEOMEAN PGA MAPS The USGS has prepared the four new sets of maps for Chapter 22 of the 2009 NEHRP Recommended Seismic Provisions: 1. Maps of uniform-hazard (2 percent in 50-year) ground motions, 2. Maps of the risk coefficients for converting 2 percent in 50-year uniform-hazard ground motions to 1 percent in 50-year risk-targeted probabilistic ground motions, 3. Maps of deterministic ground motions (consistent with site-specific criteria of Section 21.2.2), and maps of peak ground accelerations for the evaluation of the potential for liquefaction and soil strength loss (according to Section 11.8.3). Because this would have resulted in a substantial increase in the number of maps, the BSSC Provisions Update Committee recommended that the separate maps for regions of the United States and its territories that appeared in ASCE/SEI 7-05 be consolidated (for the uniform-hazard ground motion, risk coefficient, and deterministic ground motion maps), into the single figures in Chapter 22. Thus, the total number of map figures (11) in these Provisions (2009) is less than that in ASCE/SEI 7- 05 (i.e., 20). Because the consolidated map figures are relatively small and difficult to read, the USGS website that automates use of the maps and formulas will be especially useful (http://earthquake.usgs.gov.designmaps/usapp). As described in the commentary to Chapter 21 and below, the uniform-hazard and deterministic ground motion maps in Chapter 22 of these Provisions (2009) represent response in the maximum direction. The USGS has developed these maps based on "geomean" ground motions (the product of hazard assessment using modern ground motion attenuation functions), adjusted using constant factors that transform geomean response to maximum direction response. The same factors (i.e., 1.1 at short-periods and 1.3 at a period of 1 second) are used for all seismic regions (i.e., both the central and eastern United States or CEUS and the western United States or WUS) and for both probabilistic and deterministic ground motions. In contrast, the peak ground acceleration maps in Chapter 22 represent geomean ground motions, as described below. Furthermore, the peak ground acceleration maps represent the lesser of uniform-hazard (2 percent in 50-year) and deterministic peak ground accelerations, without consideration of corresponding risk coefficients. Uniform-Hazard (2 Percent in 50-Year) Ground Motion Maps The uniform-hazard maps in Chapter 22 of these Provisions (2009) are based on the 2008 USGS National Seismic Hazard Maps (http://earthquake.usgs.gov/hazmaps); however, since the ground motion values on the uniform-hazard maps are for the maximum direction of acceleration (as explained above), they are different from the “geomean” USGS maps. The 0.2- second and 1-second spectral response acceleration uniform-hazard maps are different by a factor of 1.1 and 1.3 from the respective USGS maps. Development of the USGS maps is documented in Petersen et al. (2008). Risk Coefficient Maps Development of risk coefficients and related work by the USGS is documented by Luco et al. (2007). The risk coefficient maps indicate that, in general, risk-targeted probabilistic ground motions (based on 1 percent in 50-year collapse risk) would moderately decrease the uniform-hazard ground motions (based on 2 percent in 50-year hazard) in high-hazard areas of the CEUS and the coastal region of Oregon (by as much as 30 percent) and either slightly increase or decrease the uniformhazard ground motions in the WUS and remaining areas of the maps (by less than 15 percent). These changes do not affect calculation of deterministic ground motions, which often govern in high seismic areas. Deterministic Ground Motion Maps The deterministic maps in Chapter 22 of the Provisions represent the greater of 84th percentile (maximum direction) response and the “water level” values described in the next paragraph. The USGS has developed these maps based on median "geomean" ground motions (the product of hazard assessment using modern ground motion attenuation functions) adjusted using factors that transform median geomean-direction response to 84th percentile maximum-direction response. The same factors (i.e., 1.1 x 1.8 at short-periods and 1.3 x 1.8 at a period of 1 second) are used for all seismic regions (i.e., both the CEUS and WUS regions). As defined in ASCE/SEI 7-05 Section 21.2.2, the deterministic spectral response accelerations (for Site Class B) shall not be taken as lower than 1.5g for the short periods and 0.6g for the 1-second period; hence, the ground motions on the deterministic maps (Figures 22-3 and 22-4) are no lower than these values. Otherwise the ground motions on the deterministic maps are 180 percent (as opposed 150 percent in ASCE/SEI 7-05) of median spectral response accelerations, for reasons explained above in the commentary to Chapter 21. Like the uniform-hazard maps described above, the deterministic maps represent the spectral response acceleration in the maximum direction. Peak Ground Acceleration Maps Unlike the uniform-hazard and deterministic ground motion maps described above, the peak ground acceleration maps in Chapter 22 of the Provisions represent geometric mean ground motions (not response in the maximum direction). Despite representing geometric mean ground motions, the peak ground acceleration maps are different from the 2008 USGS National Seismic Hazard Maps (http://earthquake.usgs.gov/hazmaps) upon which they are based. This is because they represent the lesser of uniform-hazard (2 percent in 50-year hazard) and deterministic peak ground accelerations. Development of the uniform-hazard peak ground accelerations is documented in Petersen et al. (2008). The deterministic peak ground accelerations are calculated as the greater of 180 percent of median ground motions and a water level of 0.6g, Note that risk coefficients are not included in the development of the peak ground acceleration maps, which is why they are referred to as “maximum considered earthquake geometric mean peak ground acceleration” maps without the “risk-targeted” prefix. REFERENCES American Society of Civil Engineers. 2006. Minimum Design Loads for Buildings and Other Structures, ASCE/SEI 7-05. ASCE, Reston, Virginia. Luco, N., B. R. Ellingwood, R. O. Hamburger, J. D. Hooper, J. K. Kimball and C. A. Kircher. 2007. “Risk-Targeted versus Current Seismic Design Maps for the Conterminous United States,” in Proceedings of the SEAOC 76th Annual Convention. Structural Engineers Association of California, Sacramento, California. Petersen, M. D., A. D. Frankel, S. C. Harmsen, C. S. Mueller, K. M. Haller, R. L. Wheeler, R. L. Wesson, Y. Zeng, O. S. Boyd, D. M. Perkins, N. Luco, E. H. Field, C. J. Wills, and K. S. Rukstales. 2008. Documentation for the 2008 Update of the United States National Seismic Hazard Maps, USGS Open File Report 2008-1128. USGS, Golden, Colorado. Page intentionally left blank. Modifications to Chapter 23, Seismic Design Reference Documents SECTION 23.1, CONSENSUS STANDARDS AND OTHER REFERENCE DOCUMENTS Add the following entries: ASCE 41 Supplement 1, Section 3.3.3 Seismic Rehabilitation of Existing Buildings, 2007 ANSI/AISI S110 Sections 14.1.1, 14.1.2, 14.1.3, Table 12.2-1 Standard for Seismic Design of Cold-Formed Steel Structural Systems – Special Bolted Moment Frames, 2007. ANSI/RMI MH 16.1 Section 15.5.3 Specification for the Design, Testing, and Utilization of Industrial Steel Storage Racks, 2008 Revise the following entries to read as indicated: ACI 318 Sections 14.2.2, 14.2.2.1, 14.2.2.2, 14.2.2.3, 14.2.2.4, 14.2.2.5, 14.2.2.6, 14.2.2.7, 14.2.2.8, 14.2.2.9, 14.2.3, 14.2.3.1.1, 14.2.3.2.1, 14.2.3.2.2, 14.2.3.2.3, 14.2.3.2.5, 14.2.3.2. Building Code Requirements for Structural Concrete, 2008. NFPA 13 Sections 13.6.5.1, 13.6.8, 13.6.8.2, 13.6.8.4 Standard for the Installation of Sprinkler Systems, 2007 Delete the following entry: RMI Rack Manufacturers Institute 8720 Red Oak Boulevard Suite 201 Charlotte, NC 28217 RMI Section 15.5.3 Specification for the Design, Testing, and Utilization of Industrial Steel Storage Racks 1997, reaffirmed 2002 Page intentionally left blank. New Chapter 23, Vertical Ground Motions for Seismic Design Add the following new Chapter 23 and renumber the existing ASCE/SEI 7-05 Chapter 23 as Chapter 24: Chapter 23 VERTICAL GROUND MOTIONS FOR SEISMIC DESIGN 23.1 DESIGN VERTICAL RESPONSE SPECTRUM. Where a design vertical response spectrum is required by these Provisions and site-specific procedures are not used, the design vertical response spectral acceleration, Sav, (in g – gravity unit) shall be developed as follows: 1. For vertical periods less than or equal to 0.025 second, Sav shall be determined in accordance with Equation 23.1-1 as follows: Sav = 0.3CVSDS (23.1-1) 2. For vertical periods greater than 0.025 second and less than or equal to 0.05 second, Sav shall be determined in accordance with Equation 23.1-2 as follows: Sav = 20CVSDS(TV - 0.025)+0.3CVSDS (23.1-2) 3. For vertical periods greater than 0.05 second and less than or equal to 0.15 second, Sav shall be determined in accordance with Equation 23.1-3 as follows: Sav = 0.8CVSDS (23.1-3) 4. For vertical periods greater than 0.15 second and less than or equal to 2.0 seconds, Sav shall be determined in accordance with Equation 23.1-4 as follows: (23.1-4) where CV is defined in terms of SS in Table 23.1-1, SDS = the design spectral response acceleration parameter at short periods, and TV = the vertical period of vibration. Table 23.1-1 Values of Vertical Coefficient CV Equation 0.75 0.8 0.15 av V DS V S C S T . . = . . . . MCER Spectral Response Parameter at Short Periods a Site Class A, B Site Class C Site Class D, E, F Ss = 2.0 0.9 1.3 1.5 SS = 1.0 0.9 1.1 1.3 SS = 0.6 0.9 1.0 1.1 SS = 0.3 0.8 0.8 0.9 SS = 0.2 0.7 0.7 0.7 a Use straight-line interpolation for intermediate values of SS . Sav shall not be less than one-half (1/2) of the corresponding Sa for horizontal components determined in accordance with the general or site-specific procedures of Section 11.4 or Chapter 21, respectively. For vertical periods greater than 2.0 seconds, Sav shall be developed from a site-specific procedure; however, the resulting ordinate of Sav shall not be less than one-half (1/2) of the corresponding Sa for horizontal components determined in accordance with the general or site-specific procedures of Section 11.4 or Chapter 21, respectively. In lieu of using the above procedure, a site-specific study may be performed to obtain Sav at vertical periods less than or equal to 2.0 seconds, but the value so determined shall not be less than 80 percent of the Sav value determined from Equations 23.1-1 through 23.1-4. 23.2 MCER VERTICAL RESPONSE SPECTRUM. The MCER vertical response spectral acceleration shall be 150 percent of the Sav determined in Section 23.1. Commentary to New Chapter 23 Chapter 23 Commentary VERTICAL GROUND MOTIONS FOR SEISMIC DESIGN C23.1 DESIGN VERTICAL RESPONSE SPECTRUM General. ASCE/SEI 7-05 and the earlier editions of the Provisions use the term 0.2SDSD to reflect the effects of vertical ground motion. Where a more explicit consideration of vertical ground motion effects is advised—as for certain tanks, materials storage facilities, and electric power generation facilities—the requirements of this chapter may be applied. Historically, the amplitude of vertical ground motion has been inferred to be two-thirds (2/3) the amplitude of the horizontal ground motion. However, studies of horizontal and vertical ground motions over the past 25 years have shown that such a simple approach is not valid in many situations (e.g., Bozorgnia and Campbell, 2004, and references therein) for the following main reasons: (a) vertical ground motion has a larger proportion of short-period (high-frequency) spectral content than horizontal ground motion and this difference increases with decreasing soil stiffness and (b) vertical ground motion attenuates at a higher rate than horizontal ground motion and this difference increases with decreasing distance from the earthquake. The observed differences in the spectral content and attenuation rate of vertical and horizontal ground motion lead to the following observations regarding the vertical/horizontal (V/H) spectral ratio (Bozorgnia and Campbell, 2004): 1. The V/H spectral ratio is relatively sensitive to spectral period, distance from the earthquake, local site conditions, and earthquake magnitude (but only for relatively soft sites) and relatively insensitive to earthquake mechanism and sediment depth; 2. The V/H spectral ratio has a distinct peak at short periods that generally exceeds 2/3 in the near-source region of an earthquake; and 3. The V/H spectral ratio is generally less than 2/3 at mid-to-long periods. Therefore, depending on the period, the distance to the fault, and the local site conditions of interest, use of the traditional 2/3V/H spectral ratio can result in either an underestimation or an overestimation of the expected vertical ground motions. The procedure for defining the design vertical response spectrum in the Provisions is based on the studies of horizontal and vertical ground motions conducted by Campbell and Bozorgnia (2003) and Bozorgnia and Campbell (2004). These procedures are also generally compatible with the general observations of Abrahamson and Silva (1997) and Silva (1997) and the proposed design procedures of Elnashai (1997). General Design Procedure. In order to be consistent with the shape of the horizontal design response spectrum, the vertical design response spectrum has four regions defined by the vertical period of vibration (Tv). Based on the study of Bozorgnia and Campbell (2004), the periods that define these regions are approximately constant with respect to the magnitude of the earthquake, the distance from the earthquake, and the local site conditions. In this respect, the shape of the vertical response spectrum is simpler than that of the horizontal response spectrum. The equations that are used to define the design vertical response spectrum are based on three observations made by Bozorgnia and Campbell (2004): 1. The short-period part of the 5 percent damped vertical response spectrum is controlled by the spectral acceleration at Tv = 0.1 second; 2. The mid-period part of the vertical response spectrum is controlled by a spectral acceleration that decays as the inverse of the 0.75 power of the vertical period of vibration (Tv -0.75); and 3. The short-period part of the V/H spectral ratio is a function of the local site conditions, the distance from the earthquake (for sites located within about 60 km of the fault), and the earthquake magnitude (for soft sites). The Provisions do not include seismic design maps for the vertical spectral acceleration at Tv = 0.1 second and do not preserve any information on the earthquake magnitudes or the source-to-site distances that contribute to the horizontal spectral accelerations that are mapped. Therefore, the general procedure recommended by Bozorgnia and Campbell (2004) was modified to use only those horizontal spectral accelerations that are available from the seismic design maps, as follows: 1. Estimate the vertical spectral acceleration at Tv = 0.1 second from the ratio of this spectral acceleration to the horizontal spectral acceleration at T = 0.2 second for the Site Class BC boundary (i.e., the boundary between Site Classes B and C ( m/sec), the reference site condition for the 2008 U.S. Geological Survey National Seismic Hazard Maps). For earthquakes and distances for which the vertical spectrum might be of engineering interest (magnitudes greater than 6.5 and distances less than 60 km), this ratio is approximately 0.8 for all site conditions (Campbell and Bozorgnia, 2003). Equation 760 s v = 2. Estimate the horizontal spectral acceleration at T = 0.2 second from the Next Generation Attenuation (NGA) relationship of Campbell and Bozorgnia (2008) for magnitudes greater than 6.5 and distances ranging between 1 and 60 km for the Site Class BC boundary ( m/sec). The relationship of Campbell and Bozorgnia (2008), rather than that of Campbell and Bozorgnia (2003), was used for this purpose in order to be consistent with the development of the 2008 U.S. Geological Survey National Seismic Hazard Maps, which use the NGA attenuation relationships to estimate horizontal ground motions in the western United States. Similar results were found for the other two NGA relationships that were used to develop the seismic hazard and design maps (Boore and Atkinson, 2008; Chiou and Youngs, 2008). Equation 760 s v = 3. Use the dependence between the horizontal spectral acceleration at T = 0.2 second and source-site distance estimated in Item 2 and the relationship between the V/H spectral ratio, source-site distance, and local site conditions in Bozorgnia and Campbell (2004) to derive a relationship between the vertical spectral acceleration and the mapped MCER spectral response acceleration parameter at short periods, SS. 4. Use the dependence between the vertical spectral acceleration and the mapped MCER spectral response acceleration parameter at short periods, SS, in Item 3 to derive a vertical coefficient, Cv, that when multiplied by 0.8 and the design horizontal response acceleration at short periods, SDS, results in an estimate of the design vertical spectral acceleration at Tv = 0.1 second. Detailed Design Procedure. The following description of the detailed design procedure listed in Section 23.1 refers to the illustrated design vertical response spectrum in Figure C23.1-1. Vertical periods less than or equal to 0.025 second. Equation 23.1-1 defines that part of the design vertical response spectrum that is controlled by the vertical peak ground acceleration. The 0.3 factor was approximated by dividing the 0.8 factor that represents the ratio between the vertical spectral acceleration at Tv = 0.1 second and the horizontal spectral acceleration at T = 0.2 second by 2.5, the factor that represents the ratio between the design horizontal spectral acceleration at T = 0.2 second, SDS, and the zero-period acceleration used in the development of the design horizontal response spectrum. The vertical coefficient, Cv, in Table 23.1-1 accounts for the dependence of the vertical spectral acceleration on the amplitude of the horizontal spectral acceleration and the site dependence of the V/H spectral ratio as determined in Items 3 and 4 above. The factors are applied to SDS rather than to SS because SDS already includes the effects of local site conditions and the 2/3 factor that is required to reduce the horizontal spectral acceleration from its MCER value to its design value. Vertical periods greater than 0.025 second and less than or equal to 0.05 second. Equation 23.1-2 defines that part of the design vertical response spectrum that represents the linear transition from the part of the spectrum that is controlled by the vertical peak ground acceleration and the part of the spectrum that is controlled by the dynamically amplified short-period spectral plateau. The factor of 20 is the factor that is required to make this transition continuous and piecewise linear between these two adjacent parts of the spectrum. Vertical periods greater than 0.05 second and less than or equal to 0.15 second. Equation 23.1-3 defines that part of the design vertical response spectrum that represents the dynamically amplified short-period spectral plateau. Vertical periods greater than 0.15 second and less than or equal to 2.0 seconds. Equation 23.1-4 defines that part of the design vertical response spectrum that decays with the inverse of the vertical period of vibration raised to the 0.75 power. Limits Imposed on Sav. Two limits are imposed on the design vertical response spectrum defined by Equations 23.1-1 through 23.1-4 and illustrated in Figure 23.1-1. The first limit restricts the vertical period of vibration to be no larger than 2 seconds. This limit accounts for the fact that such large vertical periods are rare (structures are inherently stiff in the vertical direction) and that the vertical spectrum might decay differently with period at longer periods. There is an allowance for developing a site-specific design vertical response spectrum if this limit is exceeded (see Section 11.4 or Chapter 21 for guidance on applying site-specific methods). The second limit restricts the design vertical response spectrum to be no less than 50 percent of the design horizontal response spectrum. This limit accounts for the fact that a V/H spectral ratio of onehalf (1/2) is a reasonable, but somewhat conservative, lower bound over the period range of interest, based on the results of Campbell and Bozorgnia (2003) and Bozorgnia and Campbell (2004). Figure C23.1-1 An Illustrative example of the design vertical response spectrum. 0.5 1.0 1.5 2.0 Vertical Period, Tv (sec) Vertical Spectral Acceleration 0.15 0.05 0.3 CV SDS 0.8 CV SDS 0.8 CV SDS (0.15/TV )0.75 0.025 Figure C23.1-1 Illustrative example of the design vertical response spectrum. REFERENCES Abrahamson, N. A., and W. J. Silva. 1997. “Empirical Response Spectral Attenuation Relations for Shallow Crustal Earthquakes,” Seismological Research Letters, 68:94–127. Boore, D. M., and G. M. Atkinson. 2008. “Ground-Motion Prediction Equations for the Average Horizontal Component of PGA, PGV, and 5%-Damped PSA at Spectral Periods Between 0.01 S and 10.0 s”, Earthquake Spectra, 24:99-138. Bozorgnia, Y., and K. W. Campbell. 2004. “The Vertical-to-Horizontal Response Spectral Ratio and Tentative Procedures for Developing Simplified V/H and Vertical Design Spectra,” Journal of Earthquake Engineering, 8:175-207. Campbell, K. W., and Y. Bozorgnia. 2008. “NGA Ground Motion Model for the Geometric Mean Horizontal Component of PGA, PGV, PGD and 5% Damped Linear Elastic Response Spectra for Periods Ranging from 0.01 to 10 s,” Earthquake Spectra, 24:139-171. Campbell, K. W., and Y. Bozorgnia. 2003. “Updated Near-source Ground Motion (Attenuation) Relations for the Horizontal and Vertical Components of Peak Ground Acceleration and Acceleration Response Spectra,” Bulletin of the Seismological Society of America, 93:314–331. Chiou, B. S.-J., and R. R. Youngs. 2008. “An NGA Model for the Average Horizontal Component of Peak Ground Motion and Response Spectra,” Earthquake Spectra, 24:173-215. Elnashai, A. S. 1997. “Seismic Design with Vertical Earthquake Motion,” in Seismic Design for the Next Generation of Codes, edited by P. Fajfar and H. Krawinkler. Balkema, Rotterdam, p. 91–100. Silva, W. 1997. “Characteristics of Vertical Strong Ground Motions for Applications to Engineering Design,” in FHWA/NCEER Workshop on the National Representation of Seismic Ground Motion for New and Existing Highway Facilities, Technical Report NCEER-97-0010. National Center for Earthquake Engineering Research, Buffalo, New York. 2009 NEHRP RECOMMENDED SEISMIC PROVISIONS FOR NEW BUILDINGS AND OTHER STRUCTURES: PART 2, COMMENTARY TO ASCE/SEI 7-05 This part of the 2009 NEHRP Recommended Seismic Provisions for New Buildings and Other Structures presents commentary to ASCE/SEI 7-05 utilizing the chapter and section numbers of that standard. Commentary to the modifications of the standard that appear in Part 1 of this Provisions volume is presented at the end of each chapter of modifications and can be used to replace or add to this Part 2 Commentary (e.g., this Part 2 Commentary addresses the maps that appear in ASCE/SEI 7-05, not the new risk-targeted maps and procedures presented in Part 1 of this volume). This commentary is intended primarily for design professionals and members of the codes- and standards-development community. However, an understanding of the basis for the seismic regulations contained in the nation’s building codes and standards is important to many outside this technical community including elected officials and other decision makers responsible for aspects of the built environment, the financial and insurance communities, and individual business owners and other citizens. These individuals and others who do not have in-depth technical knowledge may find a complementary report that presents a brief overview of the 2009 Provisions of interest. This overview is published as FEMA P-749, Concepts of Earthquake-resistant Design: An Introduction to the NEHRP Recommended Seismic Provisions for New Buildings and Other Structures. Page intentionally left blank. COMMENTARY TO CHAPTER 11, SEISMIC DESIGN CRITERIA C11.1 GENERAL C11.1.1 Purpose. When prescribed wind loading governs the stress or drift design, the resisting system still must conform to the special requirements for seismic-force-resisting systems. This is required in order to resist, in a ductile manner, potential seismic loads in excess of the prescribed wind loads. A proper, continuous load path is an obvious design requirement, but experience has shown that it often is overlooked and that significant damage and collapse can result. The basis for this design requirement is two-fold: 1. To ensure that the design has fully identified the seismic-force-resisting system and its appropriate design level and 2. To ensure that the design basis is fully identified for the purpose of future modifications or changes in the structure. Detailed requirements for analyzing and designing this load path are given in the appropriate design and materials chapters. C11.1.2 Scope. The scope statement establishes in general terms the applicability of ASCE/SEI 7-05. Certain structures are exempt for the following reasons: Exemption 1 – Detached one- and two-family dwellings in Seismic Design Categories A, B, and C, along with those located where Ss < 0.4g, are exempt because they represent low seismic risks. Exemption 2 – Structures constructed using the conventional light-frame construction requirements in Section 12.5 are deemed capable of resisting the anticipated seismic forces. While specific elements of conventional light-frame construction may be calculated to be overstressed, typically there is a great deal of redundancy and uncounted resistance in such structures. Detached one- and two-story wood-frame dwellings generally have performed well even in regions of higher seismicity. Section 12.5 adequately provides the level of safety required for such dwellings without imposing any additional requirements. Exemption 3 – Agricultural storage structures generally are exempt from most code requirements because of the exceptionally low risk to human life involved. Exemption 4 – Bridges, transmission towers, nuclear reactors, and other structures with special configuration and uses are not covered. The regulations for buildings and building-like structures presented in this document do not adequately address the design and performance of such special structures. ASCE/SEI 7-05 is not retroactive and usually applies to existing structures only when there is an addition, change of use, or alteration. Minimum acceptable seismic resistance of existing buildings is a policy issue normally set by the authority having jurisdiction. Appendix 11B of the standard contains rules of application for basic conditions. ASCE/SEI 31, Seismic Evaluation of Buildings, and ASCE/SEI 41, Seismic Rehabilitation of Existing Buildings, provide technical guidance but do not contain policy recommendations. A chapter in the International Building Code (IBC) applies to alteration, repair, addition, and change of occupancy of existing buildings, and the International Code Council maintains the International Existing Building Code (IEBC) and an associated commentary. C11.1.4 Alternate Materials and Alternate Means and Methods of Construction. It is not possible for a design standard to provide criteria for the use of all possible materials and their combinations and methods of construction, either existing or anticipated. While not citing specific materials or methods of construction currently available that require approval, this section serves to emphasize that the evaluation and approval of alternate materials and methods require a recognized and accepted approval system. The requirements for materials and methods of construction contained within the document represent the judgment of the best use of the materials and methods based on well-established expertise and historical seismic performance. It is important that any replacement or substitute be evaluated with an understanding of all the ramifications of performance, strength, and durability implied by the standard. It also is recognized that until needed standards and agencies are created, authorities having jurisdiction need to operate on the basis of the best evidence available to substantiate any application for alternates. If accepted standards are lacking, it is strongly recommended that applications be supported by extensive reliable data obtained from tests simulating, as closely as is practically feasible, the actual load and deformation conditions to which the material is expected to be subjected during the service life of the structure. These conditions, when applicable, should include several cycles of full reversals of loads and deformations in the inelastic range. C11.4 SEISMIC GROUND MOTION VALUES1 The approach adopted in Section 11.4 is intended to provide for a uniform margin against collapse at the design ground motion. In order to accomplish this objective, ground motion hazards are defined in terms of maximum considered earthquake (MCE) ground motions, which are based on a set of rules that depend on the seismic hazard of a region. Design ground motions are based on a lower bound estimate of the margin against collapse inherent in structures designed to the seismic provisions in the standard. This lower bound was judged, based on experience, to correspond to a factor of about 1.5 in ground motion. Consequently, the design earthquake ground motion was selected at a ground shaking level that is 1/1.5 (or 2/3) of the MCE ground motion. For most regions of the nation, the MCE ground motion is defined with a uniform probability of exceedance of 2 percent in 50 years (return period of about 2500 years). While stronger shaking than this could occur, it was judged that it would be economically impractical to design for such very rare ground motions and that the selection of the 2 percent probability of exceedance in 50 years as the MCE ground motion would result in acceptable levels of seismic safety. In regions of high seismicity, such as in many areas of California, the seismic hazard is typically controlled by largemagnitude events occurring on a limited number of well-defined fault systems. Probabilistic ground motions calculated at a 2 percent probability of exceedance in 50 years can be much larger than deterministic ground motions computed based on the characteristic magnitudes of earthquakes on these known active faults. These probabilistic motions are greater if these major active faults produce characteristic earthquakes every few hundred years. For these regions, it is considered more appropriate to determine MCE ground motions directly by deterministic methods based on the characteristic earthquakes of these defined faults. In order to provide an appropriate level of conservatism in the design process when the deterministic approach is used to calculate MCE ground motion, the median ground motion estimated for the characteristic event is multiplied by 1.5. 1 Note that this section focuses on the methods and design procedures of ASCE/SEI 7-05 and the 2003 edition of the Provisions; commentary on the new risk-targeted maps and design procedures is presented in Part 1 of this volume following the modifications to ASCE 7 Section 11.4 and Chapter 22. C11.4.1 Mapped Acceleration Parameters. In the general procedure, these motions are computed from mapped values of the spectral response acceleration at short periods, S http://earthquake.usgs.gov/designmaps. S S , and at 1 second, S1 , for Class B sites. These Ss and S1 values may be obtained directly from Figures 22-1 through 22-14 (in Chapter 22). Development of these maps is explained in detail in Appendix A of the Part 2 – Commentary volume of the 2003 NEHRP Recommended Provisions. The 2003 Ss and S1 values also can be obtained from the U.S. Geological Survey (USGS) website: S is the mapped value of the 5-percent-damped MCE spectral response acceleration for short-period structures founded on Site Class B (firm rock) sites. The short-period acceleration has been determined at a period of 0.2 second because it was concluded that 0.2 second was reasonably representative of the shortest effective period of buildings and structures that are designed using the standard, considering the effects of soil compliance, foundation rocking, and other factors typically neglected in structural analysis. Similarly, S1 is the mapped value of the 5-percent-damped MCE spectral response acceleration at a period of 1 second on Site Class B. The spectral response acceleration at periods other than 1 second typically can be derived from the acceleration at 1 second. Consequently, for MCE ground shaking on Site Class B sites, these two response acceleration parameters, SS and S1, are sufficient to define an entire response spectrum for the period range of importance for most buildings and structures. C11.4.3 and C11.4.4 Site Coefficients and Adjusted Acceleration Parameters. Using the general procedure to obtain acceleration response parameters that are appropriate for sites with a classification other than Site Class B, the SS and S1 values must be modified as indicated in Section 11.4.3. This modification is performed using two coefficients, Fa and Fv, that respectively scale the SS and S1 values determined for Site Class B to values appropriate for other site classes. The MCE spectral response accelerations adjusted for site class are designated SMS and SM1, respectively, for short-period and 1-secondperiod response. As described above, structural design in ASCE/SEI 7-05 is performed for earthquake demands that are 2/3 of the MCE response spectra. As set forth in Section 11.4.4, two additional parameters, SDS and SD1, are used to define the acceleration response spectrum for this design level event. These parameters are 2/3 of the respective SMS and SM1 values and define a design response spectrum for sites of any characteristics and for natural periods of vibration less than the transition period, TL. Values of SMS, SM1, SDS, and SD1 can also be obtained from the USGS website cited above. The site coefficients, Fa and Fv, presented respectively in Tables 11.4-1 and 11.4-2 for the various site classes are based on the results of empirical analyses of strong-motion data and analytical studies of site response. The amount of ground-motion amplification by a soil deposit relative to bedrock depends on the wave-propagation characteristics of the soil, which can be estimated from measurements or inferences of shear-wave velocity and in turn the shear modulus for the materials as a function of the level of shaking. In general, softer soils with lower shear-wave velocities exhibit greater amplifications than stiffer soils with higher shear-wave velocities. Increased levels of ground shaking result in increased soil stress-strain nonlinearity and increased soil damping which, in general, reduces the amplification, especially for shorter periods. Furthermore, for soil deposits of sufficient thickness, soil amplification is generally greater at longer periods than at shorter periods. An extensive discussion of the development of the Fa and Fv site coefficients is presented by Dobry, et al. (2000). Since the development of these coefficients and the development of a community consensus regarding their values in 1992, earthquake events have provided additional strong-motion data from which to infer site amplifications. Analyses conducted on the basis of these more recent data are reported by a number of researchers including Crouse and McGuire (1996), Dobry et al. (1999), Silva et al. (2000), Joyner and Boore (2000), Field (2000), Steidl (2000), Rodriquez-Marek et al. (2001), Borcherdt (2002), and Stewart et al. (2003). Although the results of these studies vary, the site amplification factors are generally consistent with those in Tables 11.4-1 and 11.4-2. C11.4.5 Design Response Spectrum. The design response spectrum (Figure 11.4-1) consists of several segments. The constant-acceleration segment covers the period band from To to Ts; response accelerations in this band are constant and equal to SDS . The constant-velocity segment covers the period band from Ts to TL, and the response accelerations in this band are proportional to 1/T with the response acceleration at 1-sec period equal to SD1. The long-period portion of the design response spectrum is defined on the basis of the parameter, TL, the period that marks the transition from the constantvelocity segment to the constant-displacement segment of the design response spectrum. Response accelerations in the constant-displacement segment, where T = TL, are proportional to 1/T 2. Values of TL are provided on maps in Figures 22-15 through 22-20. The TL maps were prepared following a two-step procedure. First, a correlation between earthquake magnitude and TL was established. Then, the modal magnitude from deaggregation of the ground-motion seismic hazard at a 2-second period (1- second period for Hawaii) was mapped. Details of the procedure and the rational for it are found in Crouse et al. (2006). C11.4.7 Site-Specific Ground Motion Procedures. The objective of a site-specific ground-motion analysis is to determine ground motions for local seismic and site conditions with higher confidence than is possible using the general procedure of Sections 11.4. Near-source effects on horizontal response spectra for periods of vibration greater than approximately 0.5 second include directivity, which increases ground motions for fault rupture propagating toward the site, and directionality, which increases ground motions normal (perpendicular) to the strike of the fault. These effects are discussed in Somerville et al. (1997) and Abrahamson (2000). C11.5 IMPORTANCE FACTOR AND OCCUPANCY CATEGORY Large earthquakes are rare events that will include severe ground motions. Such events are expected to result in damage to structures even if they were designed and built in accordance with the minimum requirements of the standard. The consequence of structural damage or failure is not the same for the various types of structures located within a given community. Serious damage to certain classes of structures, such as critical facilities (e.g., hospitals), will disproportionally affect a community. The fundamental purpose of this section and subsequent requirements that depend on this section is to improve the ability of a community to recover from a damaging earthquake by tailoring the seismic protection requirements to the relative importance of that structure. That purpose is achieved by requiring better performance of those structures that: 1. Are necessary to response and recovery efforts immediately following an earthquake, 2. Present the potential for catastrophic loss in the event of an earthquake, or 3. House a very large number of occupants or occupants less able to care for themselves than the average. The first basis for seismic design in the standard is that structures will have a suitably low likelihood of collapse in the very rare event defined as the maximum considered earthquake (MCE) ground motion. A second basis is that life threatening damage, primarily from failure of nonstructural elements in and on structures, will be unlikely in an unusual but less rare earthquake ground motion, which is given as the design earthquake ground motion (defined as two-thirds of the MCE). Given the occurrence of ground motion equivalent to the MCE, a population of structures built to meet these design objectives will probably still experience substantial damage in many structures, rendering these structures unfit for occupancy or use. Experience in past earthquakes around the world has demonstrated that there will be an immediate need to treat injured people, to extinguish fires and prevent conflagration, to rescue people from severely damaged or collapsed structures, and to provide sustenance to a population deprived of its normal means. Experience also has shown that these needs are best met when structures essential to response and recovery activities remain functional. The standard addresses these objectives by requiring that each structure be assigned to one of the four occupancy categories presented in Chapter 1 and by assigning an importance factor to the structure based upon that occupancy category. (The two lowest categories, Ordinary and Low Hazard, are combined for all purposes within the seismic provisions). The occupancy category is then used as one of two components in determining the Seismic Design Category (see Section C11.6) and is a primary factor in setting drift limits for building structures under the design earthquake ground motion (see Section C12.12). Figure C11.5-1 shows the combined intent of these requirements for design. The vertical scale is the likelihood of the ground motion with the MCE being the rarest considered. The horizontal scale is the level of performance of the structure and attached nonstructural components from collapse prevention at the low end to operational at the high end. (These performance levels are discussed further at other locations in the commentary.) The basic objective of collapse prevention at the MCE for ordinary structures (Occupancy Category II) is shown at the lower right by the solid triangle; protection from life-threatening damage at the design ground motion (defined by the standard as two-thirds of the MCE) is shown by the open triangle. The performance implied for higher occupancy categories is shown by square and circles. The performance anticipated for less severe ground motion is shown by dotted symbols. The three (net) classes and the numerical values assigned are far too coarse to assure the portrayed outcome for all structures, but it is judged to be adequate for the purpose given present limitations of knowledge and tools. Figure C11.5-1 Expected performance as related to occupancy category (OC) and level of ground motion. C11.5.1 Importance Factor. The importance factor is used throughout the standard in quantitative criteria for strength. In most of those quantitative criteria, the importance factor is shown as a divisor on the factor R or Rp in order to send a message to designers that the objective is to reduce damage for important structures in addition to preventing collapse in larger ground motions. The R and Rp factors adjust the computed linear elastic response to a value appropriate for design; in many structures, the largest component of that adjustment is ductility (the ability of the structure to undergo repeated cycles of inelastic strain in opposing directions). Inelastic strain damages a structure so, for a given strength demand, reducing the effective R factor (by means of the importance factor) increases the required yield strength, thus reducing ductility demand and related damage. C11.5.2 Protected Access for Category IV Structures. Those structures considered essential facilities for response and recovery efforts must be accessible to carry out their purpose. For example, if the collapse of a simple canopy at a hospital could block ambulances from the emergency room admittance area, the canopy must meet the same structural standard as the hospital. This requirement must be considered in the siting of essential facilities in densely built urban areas. Equation PERFORMANCE LEVEL Immediate Occupancy Collapse Prevention Frequent MCE Design GROUND MOTION Operational Life Safety OC IV: Essential OC III: High Occupancy OCII: Ordinary C11.6 SEISMIC DESIGN CATEGORIES Seismic design categories (SDCs) provide a means to step progressively from simple, easily performed design and construction procedures and minimums to more sophisticated, detailed, and costly requirements as both the level of seismic hazard and the consequence of failure escalate. The SDCs are used to trigger requirements that are not scalable; such requirements are either on or off. For example, the basic amplitude of ground motion for design is scalable – the quantity simply increases in a continuous fashion as one moves from a low hazard area to a high hazard area. However, a requirement to avoid weak stories is not particularly scalable. Requirements such as this create step functions. There are many such requirements in the standard, and the SDCs are used systematically to group these step functions. (Further examples include whether seismic anchorage of nonstructural items is required or not, whether particular inspections will be required or not, and height limits applied to various structural systems.) In this regard, SDCs perform one of the functions of the seismic zones used in earlier U.S. building codes and still in use throughout much of the world. However, SDCs also are dependent on a building’s occupancy and, therefore, its desired performance. Further, unlike the traditional implementation of seismic zones, the ground motions used to define the SDCs include the effects of individual site conditions on probable ground-shaking intensity. In developing the ground-shaking limits for the various Seismic Design Categories and the design requirements for each, the equivalent modified Mercalli intensity (MMI) of various shaking spectra were considered. There are now various correlations of the qualitative MMI with quantitative characterizations of ground. The reader is encouraged to consult any of a great many sources that describe the MMIs. The following list is a very coarse generalization: MMI V No real damage MMI VI Light nonstructural damage MMI VII Hazardous nonstructural damage MMI VIII Hazardous damage to susceptible structures MMI IX Hazardous damage to robust structures When the current design philosophy was adopted (the 1997 edition of the NEHRP Recommended Provisions, FEMA 302, and Commentary, FEMA 303), the upper limit for SDC A was set at roughly one-half of the lower threshold for MMI VII and the lower limit for SDC D was set at roughly the lower threshold for MMI VIII. However, the lower limit for SDC D was more consciously established by equating that design value (two-thirds of the MCE) to one-half of what had been the maximum design value in building codes over the period of 1975 to 1995. As more correlations between MMI and numerical representations of ground motion have been created, it is reasonable to make the following correlation between the MMI at MCE ground motion and the Seismic Design Category (all this discussion is for ordinary occupancies): MMI V SDC A MMI VI SDC B MMI VII SDC C MMI VIII SDC D MMI IX SDC E An important change was made to the determination of SDC when the current design philosophy was adopted. Earlier editions of the Provisions utilized the peak velocity-related acceleration, Av, to determine a building’s Seismic Performance Category. However, this coefficient does not adequately represent the damage potential of earthquakes on sites with soil conditions other than rock. Consequently, the 1997 Provisions adopted the use of response spectral acceleration parameters SDS and SD1, which include site soil effects for this purpose. Except for the lowest level of hazard (SDC A), the SDC also depends on the occupancy categories. For a given level of ground motion, the SDC is one category higher for Occupancy Category IV structures than for lower-risk structures. This has the effect of increasing the confidence that the design and construction requirements will deliver the intended performance in the extreme event. Note that the tables in the standard are at the design level, defined as two-thirds of the MCE level. Also recall that the MMIs are qualitative by their nature and that the above correlation will be more or less valid depending on which numerical correlation for MMI is used. The numerical correlations for MMI roughly double with each step so correlation between design earthquake ground motion and MMI is not as simple or convenient. In sum, at the MCE level, SDC A structures should not see motions that are normally destructive to structural systems, whereas the MCE level motions for SDC D structures can destroy vulnerable structures. The grouping of step function requirements by SDC is such that there are a few basic structural integrity requirements imposed at SDC A graduating to a suite of requirements at SDC D based upon observed performance in past earthquakes, analysis, and laboratory research. The nature of ground motions within a few kilometers of a fault can be very different from more distant motions. For example, some near fault motions will have strong velocity pulses, associated with forward rupture directivity, that tend to be highly destructive to irregular structures even if they are well detailed. For ordinary occupancies, the boundary between SDCs D and E is set to define sites likely to be close enough to a fault that these unusual ground motions may be present. Note that this boundary is defined in terms of mapped bedrock outcrop motions affecting response at 1 second, not site adjusted values, in order to better discriminate between sites near and far from faults. Short-period response is not normally as affected as the longer period response. The additional design criteria imposed on structures in SDCs E and F specifically are intended to provide acceptable performance under these very intense near-fault ground motions. For most buildings, the SDC is determined without consideration of the building’s period. Structures are assigned to a SDC based on the more severe condition determined from 1-second acceleration and short-period acceleration. This is done for several reasons. Perhaps the most important of these is that it is often difficult to estimate precisely the period of a structure using default procedures contained in the standard. Consider, for example, the case of rigid wall/flexible diaphragm buildings including low-rise reinforced masonry and concrete tilt-up buildings with either untopped metal deck or wood diaphragms. The formula in the standard for determining the period of vibration of such buildings is based solely on the height of the structure and the length of wall present. These formulas typically indicate very short periods for such structures, often on the order of 0.2 second or less. However, the actual dynamic behavior of these buildings often is dominated by the flexibility of the diaphragm – a factor neglected by the approximate period formula. Large buildings of this type can have actual periods on the order of 1 second or more. In order to avoid misclassifying a building’s SDC by inaccurately estimating the structural period, the standard generally requires that the more severe SDC determined on the basis of short- and longperiod shaking be used. Another reason for this requirement is a desire to simplify building regulation by requiring all buildings on a given soil profile in a particular region to be assigned to the same SDC regardless of the structural type. This has the advantage of permitting uniform regulation of structural system selection, inspection and testing requirements, seismic design requirements for nonstructural components, and similar aspects of the design process regulated on the basis of SDC, within a community. Notwithstanding the above, it is recognized that classification of a building as SDC C instead of B or D can have significant impact on the cost of construction. Therefore, the 2005 edition of the standard includes an exception permitting the classification of buildings that can reliably be classified as having short structural periods on the basis of short-period shaking alone. Local or regional jurisdictions enforcing building regulations may desire to consider the effect of the maps, typical soil conditions, and Seismic Design Categories on the practices in their jurisdictional areas. For reasons of uniformity of practice or reduction of potential errors, adopting ordinances could stipulate particular values of ground motion, particular site classes, or particular Seismic Design Categories for all or part of the area of their jurisdiction. For example: 1. An area with a historical practice of high seismic zone detailing might mandate a minimum SDC of D regardless of ground motion or site class. 2. A jurisdiction with low variation in ground motion across the area might stipulate particular values of ground motion rather than requiring use of the maps. 3. An area with unusual soils might require use of a particular Site Class unless a geotechnical investigation proves a better Site Class. C11.7 DESIGN REQUIREMENTS FOR SEISMIC DESIGN CATEGORY A Seismic Design Category A is assigned when the MCE ground motions are well known to be below those normally associated with hazardous damage. Damaging earthquakes are not unknown or impossible in such regions, however, and ground motions close to such events may be large enough to produce serious damage. Providing a minimum level of resistance reduces both the radius over which the ground motion exceeds structural capacities and resulting damage in such rare events. There are reasons beyond seismic risk for minimum levels of structural integrity. The requirements for SDC A are all minimum strengths for structural elements stated as forces at the level appropriate for direct use in the strength design load combinations. The two fundamental requirements are a minimum strength for a structural system to resist lateral forces and a minimum strength for connections of structural members. For many buildings the wind force will control the strength of the lateral-force-resisting system but, for low-rise buildings of heavy construction with large plan aspect ratio, the minimum lateral force specified here may control. Note that the requirement is for strength and not for toughness, energy dissipation capacity, or some measure of ductility. The force level is not tied to any postulated seismic ground motion. The boundary between SDCs A and B is based on a spectral response acceleration of 25 percent of gravity (MCE level) for short-period structures; clearly the 1 percent acceleration level (Equation 11.7-1) is far smaller. For ground motions below the A/B boundary, the spectral displacements generally are on the order of a few inches or less depending on period. Experience has shown that even a minimal strength is beneficial in providing resistance to small ground motions, and it is an easy provision to implement in design. The low probability of motions greater than the MCE is a factor in taking the simple approach without requiring details that would produce a ductile response. Another factor is that larger design forces are specified for connections between main elements of the lateral force load path. The minimum connection force is specified in three ways: a general minimum horizontal capacity for all connections; a special minimum for horizontal restraint of beams and trusses in line, which also includes the live load on the member; and a special minimum for horizontal restraint of concrete and masonry walls perpendicular to their plane. The 5 percent coefficient used for the first two is a simple and convenient value that provides some margin over the minimum strength of the system as a whole. The value for anchorage of concrete and masonry walls is simply scaled upward from the value of 200 pounds per linear foot traditionally used in past building codes for allowable stress design. C11.8 GEOLOGIC HAZARDS AND GEOTECHNICAL INVESTIGATION In addition to this commentary, Part 3 of the 2009 NEHRP Recommend Provisions includes additional and more detailed discussion and guidance on evaluation of geologic hazards and determination of seismic lateral pressures. C11.8.1 Site Limitation for Seismic Design Categories E and F. Because of the difficulty of designing a structure for the direct shearing displacement of fault rupture and the relatively high seismic activity of SDCs E and F, locating a structure on an active fault having the potential to cause rupture of the ground surface at the structure is prohibited. C11.8.3 Additional Geotechnical Investigation Report Requirements for Seismic Design Categories D through F. The dynamic lateral earth pressure on basement and retaining walls during earthquake ground shaking is considered to be an earthquake load, E, for use in design load combinations. This dynamic earth pressure is superimposed on the pre-existing static lateral earth pressure during ground shaking. The pre-existing static lateral earth pressure is considered to be an H load. Liquefaction potential should be evaluated for design earthquake ground motions consistent with peak ground accelerations of SDS/2.5. The occurrence and consequences of geologic hazards for MCE ground motions also should be considered when evaluating structural stability and other pertinent performance criteria. REFERENCES Abrahamson, N.A. 2000. “Effects of Rupture Directivity on Probabilistic Seismic Hazard Analysis” in Proceedings of the 6th International Conference on Seismic Zonation, Palm Springs, California. Borcherdt, R. D. 2002. “Empirical Evidence for Site Coefficients in Building-code Provisions,” Earthquake Spectra, 18(2):189-217. Crouse, C. B., and J. W. McGuire. 1996. “Site Response Studies for Purposes of Revising NEHRP Seismic Provisions,” Earthquake Spectra, 12(3). Crouse, C. B., E. V. Leyendecker, P. G. Somerville, M. Power, and W. J. Silva. 2006. “Development of Seismic Ground- Motion Criteria for the ASCE/SEI 7 Standard,” Paper 533 in Proceedings 8th U.S. National Conference on Earthquake Engineering, April 18-22, 2006, San Francisco, California. Dobry, R., R. Ramos, and M. S. Power. 1999. Site Factors and Site Categories in Seismic Codes, Technical Report MCEER-99-0010. Multidisciplinary Center for Earthquake Engineering Research. Dobry, R., R. Borcherdt, C. B. Crouse, I. M. Idriss, W. B. Joyner, G. R. Martin, M. S. Power, E. E. Rinne, and R. B. Seed. 2000. “New Site Coefficients and Site Classifications System Used in Recent Building Seismic Code Provisions,” Earthquake Spectra, 16(1):41-67. Field, E. H. 2000. “A Modified Ground Motion Attenuation Relationship for Southern California that Accounts for Detailed Site Classification and a Basin Depth Effect,” Bulletin of the Seismological Society of America, 90:S209-S221. Joyner, W. B., and D. M. Boore. 2000, “Recent Developments in Earthquake Ground Motion Estimation in Proceeding of the 6th International Conference on Seismic Zonation, Palm Springs, California. Rodriguez-Marek, A., J. D. Bray, and N. Abrahamson. 2001. “An Empirical Geotechnical Site Response Procedure,” Earthquake Spectra, 17(1):65-87. Silva, W., R. Darragh, N. Gregor, G. Martin, N. Abrahamson, and C. Kircher. 2000. Reassessment of Site Coefficients and Near-fault Factors for Building Code Provisions, Program Element II, Report 98-HQ-GR-1010 to the U.S. Geological Survey. Somerville, P. G., N. F. Smith, R. W. Graves, and N. A. Abrahamson. 1997. “Modification of Empirical Strong Ground Motion Attenuation Relations to Include the Amplitude and Duration Effects of Rupture Directivity,” Seismological Research Letters, 68:199-222. Stewart, J. P, A. H. Liu, and Y. Choi. 2003. “Amplification Factors for Spectral Acceleration in Tectonically Active Regions,” Bulletin of the Seismological Society of America, 93(1):332-352. Steidl, J. H. 2000. “Site Response in Southern California for Probabilistic Seismic Hazard Analysis, Bulletin of the Seismological Society of America, 90:S149-S169. COMMENTARY TO CHAPTER 12, SEISMIC DESIGN REQUIREMENTS FOR BUILDING STRUCTURES C12.1 STRUCTURAL DESIGN BASIS The performance expectations for structures designed in accordance with ASCE/SEI 7-05 are described in Sections C11.1 and C11.5. Structures designed in accordance with the standard are likely to have a low probability of collapse but suffer serious damage if subjected to the maximum considered earthquake (MCE) or stronger ground motion. The uncertainty in performance results from variability of both ground motion and structural characteristics. Earthquakes load structures indirectly. As the ground displaces, a structure follows and vibrates. The vibration produces structural deformations with associated strains and stresses. Computation of dynamic response to earthquake ground shaking is complex. The basic methods of analysis in the standard employ the common simplification of a response spectrum. A response spectrum for a specific earthquake ground motion approximates the maximum value of response to that ground motion for simple structures without reflecting the total time history of response. The design response spectrum specified in Section 11.4 and used in the basic methods of analysis in Chapter 12 is a smoothed and normalized approximation for many different ground motions. Although the seismic requirements of the standard are stated in terms of forces and loads, there are no external forces applied to the above-ground portion of a structure during an earthquake. The design forces are intended only as approximations to generate internal forces suitable for proportioning the strength of structural elements and for estimating the deformations (when multiplied by the deflection amplification factor, Cd) that would occur in the same structure in the event of designlevel (not MCE) ground motion. C12.1.1 Basic Requirements. Chapter 12 of the standard sets forth a set of coordinated requirements that must be used together. The basic steps in structural design for acceptable seismic resistance are as follows: 1. Select gravity- and seismic-force-resisting systems appropriate to the anticipated intensity of ground shaking. Section 12.2 sets forth limitations depending on the Seismic Design Category. 2. Lay out these systems to produce a continuous, regular, and redundant load path so that the structures act as integral units in responding to ground shaking. Section 12.3 addresses configuration and redundancy issues. 3. Analyze a mathematical model of the structure subjected to lateral seismic motions and gravity forces. Sections 12.6 and 12.7 set forth requirements for the method of analysis and for construction of the mathematical model. 4. Proportion members and connections to have adequate lateral and vertical strength and stiffness. Section 12.4 specifies how the effects of gravity and seismic loads are to be combined to establish required strengths, and Section 12.12 specifies deformation limits for buildings. One- to three-story structures with shear wall or braced frame systems of simple configuration may be eligible for design under the simplified alternative contained in Section 12.14. Any other deviations from the requirements of Chapter 12 are subject to approval and must be rigorously consistent as specified in Section 11.1.4. The baseline seismic forces for proportioning structural elements (individual members, connections, and supports) are static horizontal forces derived from a linear elastic response spectrum procedure. A basic requirement is that horizontal motion can come from any direction, with detailed requirements being provided in Section 12.5. For most structures, the effect of vertical ground motions is not analyzed specifically; it is included in an approximate fashion by adjusting the load factors for dead load up and down, as described in Section 12.4. Certain conditions requiring more detailed analysis of vertical response are defined in Chapters 13 and 15 for nonstructural components and nonbuilding structures, respectively. Higher levels of seismic analysis are permitted (and encouraged) for any structure and are required for some structures (see Section 12.6), but lower limits based on the equivalent lateral force procedures apply. The basic procedure uses response spectra that are representative of, but substantially reduced from, the anticipated ground motions. As a result, at the MCE level of ground shaking, structural elements are expected to yield, buckle, or otherwise behave inelastically. This approach has substantial historical precedent. In past earthquakes, structures with appropriately ductile, regular, continuous systems designed for reduced forces have performed acceptably. In the standard, such design forces are computed by dividing the forces that would be generated in a structure behaving linearly when subjected to the design ground motion by the response modification coefficient, R, and the design ground motion is taken as two-thirds of the MCE ground motion. The elastic deformations calculated under these reduced design forces are multiplied by the deflection amplification factor, Cd, to estimate the deformations likely to result from the design ground motion. As set forth in Sections 12.12 and 13, the amplified deformations are used to assess story drifts and to determine seismic demands on elements of the structure that are not part of the seismic-force-resisting system and on nonstructural components within structures. Where Cd is substantially less than R, the system is considered to have damping greater than the nominal 5 percent of critical damping. The seismic-force-resisting system is expected to reach significant yield for forces in excess of the design forces. Significant yield is the point where complete plastification of the most critical region of the structure (e.g., formation of a first plastic hinge in the structure) occurs, not the point where first yield occurs in any member. Figure C12.1-1 shows the lateral force versus deformation relation for a typical structure. Significant yield is shown as the lowest yield hinge on the forcedeformation diagram. With increased lateral loading, additional plastic hinges form and the resistance increases (following the solid curve) until a maximum is reached. The maximum resistance developed along the curve is substantially higher than that at first significant yield, and the margin is referred to as the overstrength capacity. The provisions of the standard contemplate a seismic-force-resisting system with redundant characteristics wherein significant structural overstrength above the level of significant yield can be obtained by plastification at other points in the structure prior to the formation of a complete mechanism. The overstrength obtained by this continued inelastic action provides the reserve strength necessary for the structure to resist the extreme motions of the actual seismic forces that may be generated by the design ground motion. The structural overstrength described above results from the development of sequential plastic hinging in a properly designed, redundant structure. Several other sources will further increase structural overstrength. First, material overstrength (i.e., actual material strengths higher than the nominal material strengths specified in the design) may increase the structural overstrength significantly. For example, a recent survey shows that the mean yield strength of A36 steel is about 30 to 40 percent higher than the minimum specified strength used in design calculations. Second, member design strengths usually incorporate a strength reduction (or resistance) factor, F, to produce a low probability of failure under design loading. Third, designers themselves introduce additional overstrength by selecting sections or specifying reinforcing patterns that exceed those required by the computations. Similar situations occur where prescriptive minimums of the standard, or of the design standards referenced from it, control the design. Finally, the design of many flexible structural systems (e.g., moment resisting frames) often is controlled by the drift rather than strength limitations of the standard with sections selected to control lateral deformations rather than to provide the specified strength. The result is that structures typically have a much higher lateral resistance than that specified as a minimum by the standard, and first significant yielding of structures may occur at lateral load levels that are 30 to 100 percent higher than the prescribed design seismic forces. If provided with adequate ductile detailing, redundancy and regularity, full yielding of structures may occur at load levels that are two to four times the prescribed design force levels. Most structural systems have some components or limit states that cannot provide reliable inelastic response or energy dissipation. Such components or limit states must be designed considering that the actual forces in the structure will be larger than those at first significant yield. The standard specifies an overstrength factor, O0, to amplify the prescribed forces for use in design of such components or limit states. This specified overstrength factor is neither an upper nor a lower bound; it is simply an approximation specified to provide a nominal degree of protection against undesirable behavior. Figure C12.1-1 illustrates the significance of design parameters contained in the standard including the response modification coefficient, R; the deflection amplification factor, Cd; and the system overstrength factor, O0. These design values, provided in Table 12.2-1, as well as the criteria for story drift and P-delta effects, have been established considering the characteristics of typical properly designed structures. The actual structural overstrength, O, often will be less than the tabulated factor, O0. This means that the required ductility, Rd, usually will exceed R/O0. If excessive “optimization” of a structural design is performed with lateral resistance provided by only a few elements, the successive yield hinge behavior depicted in Figure C12.1-1 will not be able to form, the actual overstrength (O) will be small, and use of the design parameters in the standard may not provide the intended seismic performance. The response modification coefficient, R, represents the ratio of the forces that would develop under the specified ground motion if the structure had entirely linear-elastic response to the prescribed design forces (see Figure C12.1-1). The structure must be designed so that the level of significant yield exceeds the prescribed design force. The ratio R, expressed as R = VE /VS, is always larger than 1.0; thus, all structures are designed for forces smaller than those the design ground motion would produce in a structure with completely linear-elastic response. This reduction is possible for a number of reasons. As the structure begins to yield and deform inelastically, the effective period of response of the structure lengthens which, for most structures, results in a reduction in strength demand. Furthermore, the inelastic action results in a significant amount of energy dissipation (hysteretic damping) in addition to other sources of damping present below significant yield. The combined effect, which is also known as the ductility reduction, explains why a properly designed structure with a fully Figure C12.1-1 Inelastic force-deformation curve. Lateral seismic force, V VE Lateral deformation (drift), Fully yielded strength . Design force level E Successive yield hinges Elastic response of structure . . , S drift under design forces Vy VS E S R V V = E d y R V V = y S V V O = D d S C = . . . , D design drift yielded strength (Vy in Figure C12.1-1) that is significantly lower than the elastic seismic force demand (VE in Figure C12.1- 1) can be capable of providing satisfactory performance under the design ground motion excitations. Figure C12.1-1 Inelastic force-deformation curve. The energy dissipation resulting from hysteretic behavior can be measured as the area enclosed by the force-deformation curve of the structure as it experiences several cycles of excitation. Some structures have far more energy dissipation capacity than others. The extent of energy dissipation capacity available depends largely on the amount of stiffness and strength degradation the structure undergoes as it experiences repeated cycles of inelastic deformation. Figure C12.1-2 shows representative load deformation curves for two simple substructures such as a beam-column assembly in a frame. Hysteretic curve (a) in the figure is representative of the behavior of substructures that have been detailed for ductile behavior. The substructure can maintain nearly all of its strength and stiffness over several large cycles of inelastic deformation. The resulting force-deformation “loops” are quite wide and open, resulting in a large amount of energy dissipation. Hysteretic curve (b) represents the behavior of a substructure that has not been detailed for ductile behavior. It loses stiffness rapidly under inelastic deformation, and the resulting hysteretic loops are quite pinched. Such a substructure has much less energy dissipation than that for the substructure (a) but has a greater change in response period. The structural response is determined by a combination of energy dissipation and period modification. The R values in the standard are based largely on engineering judgment of the performance of the various materials and systems in past earthquakes. The R factor for a specific project should be chosen and used with care. For example, lower values should be used for structures possessing a low degree of redundancy wherein all the plastic hinges required for the formation of a mechanism may be formed essentially simultaneously and at a force level close to the specified design strength. This situation can result in considerably more detrimental P-delta effects. Since it is difficult for individual designers to judge the extent to which R factors should be adjusted based on the inherent redundancy of their designs, Section 12.3.4 provides a coefficient, ., that is calculated based on the removal of individual seismic-force-resisting elements. C12.1.2 Member Design, Connection Design, and Deformation Limit. Given that key elements of the seismic-forceresisting system will likely yield in response to ground motions as discussed in Section C12.1.1, it might be expected that structural connections would be required to develop the strength of connected members. Although that is a logical procedure, it is not a general requirement. The actual requirement varies by system and generally is specified in the standards for design of the various structural materials cited by reference in Section 14. Good seismic design requires careful consideration of this issue. Figure C12.1-2 Typical hysteretic curves. Force Force Displacement Displacement a. Ductile hysteresis loops b. Pinched hysteresis loops Figure C12.1-2 Typical hysteretic curves. C12.1.3 Continuous Load Path and Interconnection. In effect, Section 12.1.3 calls for the seismic design to be complete and in accordance with the principles of structural mechanics. The loads must be transferred rationally from their point of origin to the final point of resistance. This should be obvious, but it often is overlooked by those inexperienced in earthquake engineering. Design consideration should be given to potentially adverse effects where there is a lack of redundancy. Given the many unknowns and uncertainties in the magnitude and characteristics of earthquake loading, in the materials and systems of construction for resisting earthquake loadings and in the methods of analysis, good earthquake engineering practice has been to provide as much redundancy as possible in the seismic-force-resisting system of buildings. Redundancy plays an important role in determining the ability of the building to resist earthquake forces. In a structural system without redundant components, every component must remain operative to preserve the integrity of the building structure. On the other hand, in a highly redundant system, one or more redundant components may fail and still leave a structural system that retains its integrity and can continue to resist lateral forces, albeit with diminished effectiveness. While a redundancy requirement is included in Section 12.3.4, overall system redundancy can be improved by making all joints of the vertical load-carrying frame moment resisting and incorporating them into the seismic-force-resisting system. These multiple points of resistance can prevent a catastrophic collapse due to distress or failure of a member or joint. (The overstrength characteristics of this type of frame are discussed in Section C12.1.1.) The designer should be particularly aware of the proper selection of R when using only one- or two-bay rigid frames in one direction for resisting seismic loads. A single, one-bay frame or a pair of such frames provides little redundancy so the designer may wish to consider a reduced R to account for a lack of redundancy if the calculated redundancy is considered to be too low. As more one-bay frames are added to the system, however, overall system redundancy increases. The increase in redundancy is a function of frame placement and total number of frames. The minimum connection forces are not intended to be applied simultaneously to the entire seismic-force-resisting system. C12.1.4 Connection to Supports. The requirement is the same as given in Section 11.7.4 for Seismic Design Category A. See Section C11.7. C12.1.5 Foundation Design. Most foundation design criteria are still stated in terms of allowable stresses, and the forces computed in the standard are all based on the strength level of response. When developing strength-based criteria for foundations, all the factors cited in Section 12.1.5 require careful consideration. Section C12.13 provides specific guidance. C12.1.6 Material Design and Detailing Requirements. The design limit state for resistance to an earthquake is unlike that for any other load within the scope of the standard. The earthquake limit state is based on overall system performance, not member performance, where repeated cycles of inelastic straining are accepted as an energy dissipating mechanism. Provisions that modify customary requirements for proportioning and detailing structural members and systems are provided to produce the desired performance. C12.2 STRUCTURAL SYSTEM SELECTION C12.2.1 Selection and Limitations. For purposes of these seismic analyses and design requirements, seismic-forceresisting systems are grouped into categories as shown in Table 12.2-1. These categories are subdivided further for various types of vertical elements used to resist seismic forces. In addition, the sections for detailing requirements are specified. Specification of R factors requires considerable judgment based on knowledge of actual earthquake performance as well as research studies. The factors in Table 12.2-1 continue to be reviewed in light of recent research results. R values for the various systems were selected considering observed performance during past earthquakes, the toughness (ability to dissipate energy without serious degradation) of the system, and the amount of damping typically present in the system when it undergoes inelastic response. FEMA P-695, Quantification of Building Seismic Performance Factors (Applied Technology Council, 2009) has been developed with the purpose of establishing and documenting a methodology for quantifying building system performance and response parameters for use in seismic design. While the response modification coefficient (R factor) is a key parameter being addressed, related design parameters such as the system overstrength factor (O0) and deflection amplification factor (Cd) also are addressed. Collectively, these terms are referred to as “Seismic Performance Factors” (SPFs). Future systems will likely derive their SPFs using this methodology and existing system SPFs also may be reviewed in light of this new procedure. Building height limits have been specified in codes and standards for over 50 years. The structural system limitations and building height limits specified in Table 12.2-1 evolved from these initial limitations and were further modified by the collective expert judgment of the PUC and the ATC-3 project team (the forerunners of the PUC). They have continued to evolve over the past 30 years based on observations and testing, but the specific values are based on subjective judgment. In a bearing wall system, major load-carrying columns are omitted and the walls carry a major portion of the gravity (dead and live) loads. The walls supply in-plane lateral stiffness and strength to resist wind and earthquake loads as well as other lateral loads. In some cases, vertical trusses are employed to augment lateral stiffness. In general, this system has comparably lower values of R than other systems due to the frequent lack of redundancy for support of vertical and horizontal loads. In a building frame system, gravity loads are carried primarily by a frame supported on columns rather than by bearing walls. Some portions of the gravity load may be carried on bearing walls, but the amount carried should represent a relatively small percentage of the floor or roof area. Lateral resistance is provided by shear walls or braced frames. Light-framed walls with shear panels are intended for use only with wood and steel building frames. Although gravity-load-resisting systems are not required to provide lateral resistance, most of them do. To the extent that the gravity-load-resisting system provides additional lateral resistance, it will enhance the building’s seismic performance capability, so long as it is capable of resisting the resulting stresses and undergoing the associated deformations. In a moment-resisting frame system, moment-resisting connections between the columns and beams provide lateral resistance. In Table 12.2-1, such frames are classified as ordinary, intermediate, or special. In high Seismic Design Categories, the anticipated ground motions are expected to produce large inelastic demands so special moment frames designed and detailed for ductile response in accordance with Chapter 14 are required. In low Seismic Design Categories, the inherent overstrength in typical structural designs is such that the anticipated inelastic demands are reduced somewhat, and less ductile systems may be employed safely. Since these less ductile ordinary framing systems do not possess as much toughness, lower R values are specified. The R, O0, and Cd values for the composite systems in Table 12.2-1 are similar to those for comparable systems of structural steel and reinforced concrete. Use of the tabulated values is allowed only when the design and detailing requirements in Section 14.3 are followed. In a dual system, a three-dimensional space frame made up of columns and beams provides primary support for gravity loads. Primary lateral resistance is supplied by shear walls or braced frames, and secondary lateral resistance is provided by a moment frame complying with the requirements of Chapter 14. Where a beam-column frame or slab-column frame lacks special detailing, it cannot act as an effective backup to a shear wall subsystem so there are no dual systems with ordinary moment frames. Instead, Table 12.2-1 permits the use of a shear wallframe interactive system with ordinary reinforced concrete moment frames and ordinary reinforced concrete shear walls. Use of this defined system, which requires compliance with Section 12.2.5.10, offers a significant advantage over a simple combination of the two constituent ordinary reinforced concrete systems. Where those systems are simply combined, Section 12.2.3.2 would require use of design parameters for an ordinary reinforced concrete moment frame. In a cantilevered column system, stability of mass at the top is provided by one or more columns with base fixity acting as a single-degree-of-freedom system. Cantilever column systems are essentially a special class of moment-resisting frame except that they do not possess the redundancy and overstrength that most moment-resisting frames derive from sequential formation of yield or plastic hinges. Where a typical moment-resisting frame must form multiple plastic hinges in members in order to develop a yield mechanism, a cantilever column system develops hinges only at the base of the columns to form a mechanism. As a result, their overstrength is limited to that provided by material overstrength and by design conservatism. It is permitted to construct cantilever column structures using any of the systems that can be used to develop moment frames including ordinary, intermediate, and special steel and concrete detailing systems as well as timber frames. The system limitations for cantilever column systems reflect the type of moment frame detailing provided but with a height limit of 35 feet. The R factor for cantilever column systems is derived from moment-resisting frame values where R is divided by O0 but is not taken as less than 1 or greater than 3. This accounts for the lack of sequential yielding in such systems. Cd is taken as equal to R, recognizing that damping is quite low in these systems and inelastic displacement of these systems will not be less than the elastic displacement. C12.2.2 Combinations of Framing Systems in Different Directions. Different systems can be utilized along each of the two orthogonal directions as long as the respective R, O0, and Cd values are used. Depending on the combination selected, it is possible that one of the two systems will limit the extent of the overall system with regard to use and height. The more restrictive of the limitation systems governs. C12.2.3 Combinations of Framing Systems in the Same Direction. C12.2.3.1 R, O0, and Cd Values for Vertical Combinations. The intent of the provision requiring us of the more stringent seismic design parameters (R, O0, and Cd) is to prevent mixed systems that could concentrate inelastic behavior in the lower stories. Exceptions to these requirements exist for conditions that do not affect the dynamic characteristics of the structure or that will not result in concentration of inelastic demand in critical areas. For the past several decades, building codes have allowed two-stage static analysis for certain structures with a vertical combination of dynamically uncoupled systems. While this approach may be used for any structure that meets the requirements, it is most often used for the design of light-framed construction built on a rigid concrete base. The design process requires that the “flexible” upper structure and “rigid” lower structure be designed separately with the reactions from the upper portion amplified by the ratio of respective R/. values. This ratio, which must be taken as no less than 1, produces demands for the “rigid” lower portion that are commensurate with its inelastic capability. C12.2.3.2 R, O0, and Cd Values for Horizontal Combinations. For nearly all conditions, the least value of R of different structural systems in the same direction must be used in design. This requirement reflects the expectation that the entire system will undergo the same deformation with its behavior controlled by the least ductile system. However, where the listed conditions are met, the R value for each independent line of resistance can be used. This exceptional condition is consistent with light-frame construction that utilizes the ground for parking with residential use above. C12.2.4 Combination of Framing Detailing Requirements. This requirement is provided so that the higher R value system has the necessary ductile detailing throughout. The intent is that details common to both systems be designed to remain functional throughout the response in order to preserve the integrity of the seismic-force-resisting system. C12.2.5 System Specific Requirements. C12.2.5.1 Dual System. The moment frame of a dual system must be capable of resisting at least 25 percent of the design seismic forces; this percentage is based on judgment. The purpose of the 25 percent frame is to provide a secondary lateral system with higher degrees of redundancy and ductility in order to improve the ability of the building to support the service loads (or at least the effect of gravity loads) after strong earthquake shaking. The primary system (walls or bracing) acting together with the moment frame must be capable of resisting all of the design seismic forces. The following analyses are required for dual systems: 1. The moment frame and shear walls or braced frames must resist the design seismic forces considering fully the force and deformation interaction of the walls or braced frames and the moment frames as a single system. This analysis must be made in accordance with the principles of structural mechanics considering the relative rigidities of the elements and torsion in the system. Deformations imposed upon members of the moment frame by their interaction with the shear walls or braced frames must be considered in this analysis. 2. The moment frame must be designed with sufficient strength to resist at least 25 percent of the design seismic forces including torsional effects. C12.2.5.2 Cantilever Column Systems. Cantilever column systems are singled out for special consideration because of their unique characteristics. These structures often have limited redundancy and overstrength and concentrate inelastic behavior at their bases. As a result, they have substantially less energy dissipation capacity than other systems. A number of apartment buildings incorporating this system experienced very severe damage and, in some cases, collapse in the 1994 Northridge earthquake. Because the ductility of columns having large axial stress is limited, cantilever column systems may not be used where column axial demands exceed 15 percent of their axial strength. Elements providing restraint at the base of cantilever columns must be designed with overstrength so that the strength of the cantilever columns is developed. C12.2.5.3 Inverted Pendulum-Type Structures. Inverted pendulum-type structures do not have unique entry in Table 12.2-1 since they can be formed from many structural systems. Inverted pendulum-type structures have more than half of their mass concentrated near the top (producing one degree of freedom in horizontal translation) and rotational compatibility of the mass with the column (producing vertical accelerations acting in opposite directions). Dynamic response amplifies this rotation; hence, the bending moment induced at the top of the column can exceed that computed using the procedures of Section 12.8. The requirement to design for a top moment that is one-half of the base moment calculated in accordance with Section 12.8 is based on analyses of inverted pendulums covering a wide range of practical conditions. C12.2.5.4 Increased Building Height Limit for Steel Braced Frames and Special Reinforced Concrete Shear Walls. The first criterion for an increased building height limit precludes extreme torsional irregularity since premature failure of one of the single walls or frames could lead to excessive inelastic torsional response. The second criterion, which is similar to the redundancy requirements, is to limit the height of systems that are too strongly dependent on any single line of walls or braced frames. The inherent torsion resulting from the distance between the center or mass and center of stiffness must be included, but accidental torsional effects are neglected for ease of implementation. C12.2.5.5 Special Moment Frames in Structures Assigned to Seismic Design Categories D through F. Special moment frames, either alone or as part of a dual system, are required to be used in Seismic Design Categories D through F where the building height exceeds 160 feet (or 240 feet for buildings that meet the provisions of Section 12.2.5.4) as indicated in Table 12.2-1. In shorter buildings where special moment frames are not required to be used, the special moment frames may be discontinued and supported on less ductile systems as long as the requirements for system combinations are followed. For the situation where special moment frames are required, they should be continuous to the foundation. In cases where the foundation is located below the building’s base, provisions for discontinuing the moment frames can be made as long as the seismic forces are properly accounted for and transferred to the supporting structure. C12.2.5.6 Single-Story Steel Ordinary and Intermediate Moment Frames in Structures Assigned to Seismic Design Category D or E. Ordinary and intermediate moment frames are less ductile than special moment frames; consequently, restrictions are placed on their use in higher Seismic Design Categories. The height limit of 65 feet and the limitations on roof and wall dead load are intended to restrict the use of such systems to metal buildings and similar one-story structures, the design of which is often controlled by wind forces, and which have generally evidenced acceptable performance in past seismic events. C12.2.5.7 Other Steel Ordinary and Intermediate Moment Frames in Structures Assigned to Seismic Design Category D or E. Compared to the limits in Section 12.2.5.6, this section imposes a stricter height limit because higher loads and additional stories are permitted. Low-rise light-frame structures that are commonly used in residential construction generally have evidenced adequate performance in past seismic events due to their light weight, abundance of lateral forceresisting elements, and general resilience. C12.2.5.8 Single-Story Steel Ordinary and Intermediate Moment Frames in Structures Assigned to Seismic Design Category F. See Section C12.2.5.6. C12.2.5.9 Other Steel Intermediate Moment Frame Limitations in Structures Assigned to Seismic Design Category F. The intent of this section is to prohibit the use of steel ordinary moment frames in light-frame construction that does not comply with Section 12.2.5.8. C12.2.5.10 Shear Wall-Frame Interactive Systems. For structures assigned to Seismic Design Category A or B (where seismic hazard is low), it is usual practice to design shear walls and frames of a shear wall-frame structure to resist lateral forces in proportion to their relative rigidities, considering interaction between the two subsystems at all levels. As discussed in Section C12.2.1, this typical approach would require use of a lower R factor than that defined for shear wall-frame interactive systems. Where the special requirements of this section are satisfied, more reliable performance is expected, justifying a higher R factor. C12.3 DIAPHRAGM FLEXIBILITY, CONFIGURATION IRREGULARITIES, AND REDUNDANCY C12.3.1 Diaphragm Flexibility. Most seismic-force-resisting systems have two distinct parts: the horizontal system that distributes lateral forces to the vertical elements and the vertical system that transmits lateral forces between the floor levels and the base of the structure. The horizontal system may consist of diaphragms or a horizontal bracing system. For the majority of buildings, diaphragms offer the most economical and positive method of resisting and distributing seismic forces in the horizontal plane. Typically, diaphragms consist of metal deck (with or without concrete), concrete slabs, and wood sheathing/decking. While most diaphragms are flat, consisting of the floors of buildings, they also may be inclined, curved, warped, or folded configurations, and most diaphragms have openings. The diaphragm stiffness relative to the stiffness of the supporting vertical seismic-force-resisting system ranges from flexible to rigid and is important to define. Provisions defining diaphragm flexibility are given in Sections 12.3.1.1 through 12.3.1.3. If a diaphragm cannot be idealized as either flexible or rigid, explicit consideration of its stiffness must be included in the analysis. The diaphragms in most buildings braced by wood light-frame shear walls are semi-rigid. Because semi-rigid diaphragm modeling is beyond the capability of available software for wood light-frame buildings, it is anticipated that this requirement will be met by evaluating force distribution using both rigid and flexible diaphragm models and taking the worst case of the two. While this is in conflict with common design practice, which typically includes only flexible diaphragm force distribution for wood light-frame buildings, it is one method of capturing the effect of the diaphragm stiffness. Further detailed discussion of diaphragms can be found in Delebi, et al. (1980) and in an Applied Technology Council report on diaphragms (1981). C12.3.1.2 Rigid Diaphragm Condition. Span length is included in the deemed-to-comply condition as an indirect measure of the flexural contribution to diaphragm stiffness. C12.3.2 Irregular and Regular Classification. The configuration of a structure can significantly affect its performance during a strong earthquake producing the ground motion contemplated in the standard. Configuration can be divided into two aspects: horizontal and vertical. Most seismic design provisions were derived for buildings having regular configurations, but earthquakes have shown repeatedly that buildings having irregular configurations suffer greater damage. This situation prevails even with good design and construction. There are several reasons for this poor behavior of irregular structures. In a regular structure, the inelastic response produced by strong ground shaking, including energy dissipation and damage, tends to be well distributed throughout the structure. However, in irregular structures, inelastic behavior can be concentrated by irregularities and result in rapid failure of structural elements in these areas. In addition, some irregularities introduce unanticipated demands into the structure, which designers frequently overlook when detailing the structural system. Finally, the elastic analysis methods typically employed in the design of structures often cannot predict the distribution of earthquake demands in an irregular structure very well, leading to inadequate design in the areas associated with the irregularity. For these reasons, the standard encourages regular configurations and prohibits gross irregularity in buildings located on sites close to major active faults where very strong ground motion and extreme inelastic demands are anticipated. C12.3.2.1 Horizontal Irregularity. A building may have a symmetric geometric shape without re-entrant corners or wings but still be classified as irregular in plan because of its distribution of mass or vertical seismic-force-resisting elements. Torsional effects in earthquakes can occur even where the centers of mass and resistance coincide. For example, ground motion waves acting on a skew with respect to the building axis can cause torsion. Cracking or yielding in an asymmetric fashion also can cause torsion. These effects also can magnify the torsion due to eccentricity between the centers of mass and resistance. Torsional irregularities are defined to address this concern. A square or rectangular building with minor re-entrant corners would still be considered regular, but large re-entrant corners creating a crucifix form would produce an irregular configuration. The response of the wings of this type of building generally differs from the response of the building as a whole, and this produces higher local forces than would be determined by application of the standard without modification. Other winged plan configurations (e.g., H-shapes) are classified as irregular even if symmetric due to the response of the wings. Significant differences in stiffness between portions of a diaphragm at a level are classified as irregularities since they may cause a change in the distribution of seismic forces to the vertical components and create torsional forces not accounted for in the distribution normally considered for a regular building. Figure C12.3-1 illustrates plan irregularities. Where there are discontinuities in the path of lateral force resistance, the structure cannot be considered to be regular. The most critical discontinuity defined is the out-of-plane offset of vertical elements of the seismic-force-resisting system. Such offsets impose vertical and lateral load effects on horizontal elements that are difficult to provide for adequately. Where vertical elements of the lateral-force-resisting system are not parallel to or symmetric about major orthogonal axes, the equivalent lateral force procedure of the standard cannot be applied appropriately so the structure is considered to be irregular. Figure C12.3-1 Building plan irregularities. Rigid Flexible Open 1. Torsional 2. Reentrant corner 3. Diaphragm discontinuity 4. Out-of-plane offset 5. Nonparallel system X Xp Yp Y Irregular: X X p > 0.15 and Y Y p > 0.15 Wall below Wall above Irregular: 1 A > 2 XY max avg Seismic force min X Y Irregular: >1.4 Extreme: max avg >1.2 = max avg 2 max avg + min open Figure C12.3-1 Building plan irregularities. C12.3.2.2 Vertical Irregularity. Vertical configuration irregularities affect the responses at the various levels and induce loads at these levels that differ significantly from the distribution assumed in the equivalent lateral force procedure given in Section 12.8. A moment-resisting frame building might be classified as having a vertical irregularity if one story is much taller than the adjoining stories and the design did not compensate for the resulting decrease in stiffness that normally would occur. Figure C12.3-2 illustrates vertical irregularities. A building is classified as irregular where the ratio of mass to stiffness in adjacent stories differs significantly. This might occur where a heavy mass (e.g., an interstitial mechanical floor) is placed at one level. Irregularity Type 3 in Table 12.3-2 applies regardless of whether the larger dimension is above or below the smaller one. Buildings with a weak-story irregularity tend to develop all of their inelastic behavior and consequent damage at the weak story, possibly leading to collapse. Section 12.3.3.2 provides an exception for Seismic Design Category B or C structures where essentially elastic response of the weak story is expected. C12.3.3 Limitations and Additional Requirements for Systems with Structural Irregularities. C12.3.3.1 Prohibited Horizontal and Vertical Irregularities in Seismic Design Categories D through F. The irregularity prohibitions of this section stem from poor performance in past earthquakes and the potential to concentrate large inelastic demands in certain portions of the structure. Even when such irregularities are permitted, they should be avoided whenever possible in all structures. Figure C12.3-2 Building vertical irregularities. K i+3 K i+2 K i+1 K i K i+3 K i+2 K i+1 K i K i+3 K i+2 K i+1 K i Mi+1 Mi Mi-1 Li+1 Li 2. Weight (Mass) 3. Geometric 4. In-Plane Discontinuity 5. Lateral Strength -- Weak Story Irregular: < 0.7K or i+1 Ki K < i (K + K + K ) 3 i+1 0.8 i+2 i+3 Extreme: < 0.6K or i+1 Ki K < i (K + K + K ) 3 i+1 0.7 i+2 i+3 1. Stiffness -- Soft Story Stri+1 Stri Labove Lbelow offset Irregular: offset > L or below offset > Labove Irregular: Str < 0.8Str Str < 0.65Str i i i+1 i+1 Extreme: Irregular: L > 1.3L i i+1 Irregular: M > 1.5M or i i+1 M i > 1.5Mi-1 Figure C12.3-2 Building vertical irregularities. C12.3.3.2 Extreme Weak Stories. Since extreme weak story irregularities are prohibited for buildings located in Seismic Design Categories D, E and F, the limitations and exceptions in this section apply only to buildings assigned to Seismic Design Category B or C. C12.3.3.3 Elements Supporting Discontinuous Walls or Frames. The purpose of this requirement is to protect the supporting elements from overload caused by overstrength of a discontinued seismic-force-resisting element. Columns, beams, slabs, or trusses may be subject to such failure so all are included in the design requirement. Overload may result from forces in either the downward or upward direction; therefore, both possibilities must be considered. Such load reversals may be especially problematic for reinforced concrete beams, weaker top laminations of glulam beams, unbraced flanges of steel beams, and steel trusses. The connection between the discontinuous element and the supporting member must be adequate to transmit the forces for which the discontinuous element is designed. For example, where the discontinuous element must be designed using the load combinations of Section 12.4.3, as is the case for a steel column in a braced frame or moment frame, its connection to the supporting member must be designed using the same load combinations. Since concrete shear walls are not required to be designed using the load combinations of Section 12.4.3, the connection between a discontinuous shear wall and the supporting member may be designed using the loads associated with the shear wall and not the load combinations with overstrength factor. C12.3.3.4 Increase in Forces Due to Irregularities for Seismic Design Categories D through F. The irregularities listed may result in loads that are distributed differently than assumed in the equivalent lateral force procedure of Section 12.8, especially as related to the interconnection of the diaphragm with vertical elements of the seismic-force-resisting system. The 25 percent increase in force is intended to account for this difference. Where the load combinations with overstrength apply, no further increase is warranted. C12.3.4 Redundancy. The desirability of redundancy, or multiple lateral-force-resisting load paths, has long been recognized. The redundancy provisions of this section reflect the belief that an excessive loss of story shear strength or development of an extreme torsional irregularity may lead to structural failure. The redundancy factor determined for each direction may differ. C12.3.4.1 Conditions Where Value of . is 1.0. This section provides a convenient list of conditions where . is 1.0. C12.3.4.2 Redundancy Factor, ., for Seismic Design Category D through F. There are two approaches to establishing a redundancy factor of 1.0. Where neither condition is satisfied, . is taken equal to 1.3. It is permitted to take . equal to 1.3 without checking either condition. The first approach is a check of the elements outlined in Table 12.3-3 for cases where the story shear exceeds 35 percent of the base shear. Parametric studies (conducted by Building Seismic Safety Council Technical Subcommittee 2 but unpublished) were used to select the 35 percent value. Those studies indicated that stories with at least 35 percent of the base shear include all stories of low-rise buildings (buildings up to 5 to 6 stories) and about 87 percent of the stories of tall buildings. The intent of this limit is to exclude penthouses and the uppermost stories from the redundancy requirements. This approach requires the removal (or loss of moment resistance) of an individual lateral-force-resisting element to determine its effect on the remaining structure. If the removal of elements, one-by-one, does not result in more than a 33 percent reduction in story strength or an extreme torsional irregularity, . may be taken as 1.0. For this evaluation, the determination of story strength requires an in-depth calculation. The intent of the check is to use a simple measure (elastic or plastic) to determine whether an individual member has a significant effect on the overall system. If the original structure has an extreme torsional irregularity to begin with, the resulting . is 1.3. Figure C12.3-3 presents a flowchart for implementing the redundancy requirements. As indicated in the table, braced frame, moment frame, shear wall, and cantilever column systems must conform to redundancy requirements. Dual systems also are included but, in most cases, are inherently redundant. Shear walls or wall piers with a height-to-length aspect ratio greater than 1.0 within any story have been included; however, the required design of collector elements and their connections for O0 times the design force may address the key issues. In order to satisfy the collector force requirements, a reasonable number of shear walls usually is required. Regardless, shear wall systems are addressed in this section so that either an adequate number of wall elements is included or the proper redundancy factor is applied. For wall piers, the height is taken as the height of the adjacent opening and generally is less than the story height. The second approach is a deemed-to-comply condition wherein the structure is regular and has a specified arrangement of seismic-force-resisting elements to qualify for . of 1.0. As part of the parametric study, simplified braced frame and moment frame systems were investigated to determine their sensitivity to the analytical redundancy criteria. This simple deemed-tocomply condition is consistent with the results of the study. Figure C12.3-3 Calculation of the redundancy factor, .. Perform linear analysis with all elements Define story X above which no more than 35% of base shear is resisted Below X is item b of Section 12.3.4.2 satisfied? Extreme torsional irregularity? Does the seismic force-resisting system comprise only shear walls or wall piers with a height-to-length ratio not greater than 1.0? Prioritize elements based on highest force or force/story shear Select an element (below X ) to remove, and perform linear analysis without that element Does the demand in any remaining element (below X ) increase by more than 50%? Does plastic mechanism analysis show that element removal decreases story strength by more than 33%? Have all likely elements been considered? Extreme torsional irregularity? . = 1.0 . = 1.3 No* No No No No Yes* Yes No Yes No Yes Yes Yes * or not considered Yes p p p p Figure C12.3-3 Calculation of the redundancy factor, .. Figure C12.3-4 Shear wall and wall pier height-to-length ratios. hwall Lwall wp hwp L Story level 2 Story level 1 Shear wall height-to length ratio = wall h Lwall Wall pier height-to length ratio = wp h Lwp = height of shear wall = height of wall pier = length of shear wall = length of wall pier wall h Lwall wp h Lwp Figure C12.3-4 Shear wall and wall pier height-to-length ratios. C12.4 SEISMIC LOAD EFFECTS AND COMBINATIONS C12.4.1 Applicability. Structural elements designated by the engineer as part of the seismic-force-resisting system typically are designed directly for seismic load effects. None of the seismic forces associated with the design base shear are formally assigned to structural elements that are not designated as part of the seismic-force-resisting system, but such elements must be designed using the load conditions of Section 12.4 and must accommodate the deformations resulting from application of seismic loads. C12.4.2 Seismic Load Effect. Section 12.4 presents the required combinations of seismic forces with other loads. The load combinations are taken from the basic load combinations of Chapter 2 of the standard with further elaboration of the seismic load effect, E. The seismic load effect includes horizontal and vertical components. For strength design, the effect of vertical seismic forces, Ev, is based on an assumed effective vertical acceleration of 0.2SDS times gravity. It may be helpful to recognize that the quantities Eh and Ev are the effects of loads, not the loads themselves. They can be tension or compression axial forces, shear, bending moments, or torsional moments. For a one-story shear wall, application of the horizontal seismic forces from V causes overturning moment and shear in the wall, both of which are Eh effects. The factor 0.2 SDS times gravity dead load corresponds to an Ev load effect that increases or decreases the axial force in the wall. In this simple example, an Eh force or moment is never added directly to an Ev force or moment because the former affects only moment and shear, while the latter affects only axial force. While the shear and moment are independent of the axial force, the capacity check of the wall may need to include all three terms (or certainly moment and axial force) simultaneously. For a diagonal brace that carries earthquake and gravity load, application of the horizontal seismic forces from V causes a brace force that has both horizontal and vertical components, and the factor 0.2 SDS times dead load produces a load effect that also affects both the horizontal and vertical components of axial force. In this case the brace force is based on Eh ± Ev. Section 12.4.2.3 presents the load combinations written using the separate horizontal and vertical load effects that constitute E. The 0.2SDS vertical acceleration effect is required to be considered in the design of all members of a structure—even those that are not part of the seismic-force-resisting system. For example, design of a gravity load-resisting prestressed concrete girder may be governed by the dead and earthquake condition, where 0.2SDSD is subtracted from the dead load. This could be the controlling condition for tension at the top of the girder. C12.4.3 Seismic Load Effect Including Overstrength Factor. Certain structural elements or actions, such as collectors in Seismic Design Categories C through F or columns supporting discontinuous walls, are required to be designed for seismic load combinations with overstrength. In such cases the seismic load effect, Em, has its horizontal component multiplied by the overstrength factor O0, as indicated in Section 12.4.3. C12.4.4 Minimum Upward Force for Horizontal Cantilevers for Seismic Design Categories D through F. In Seismic Design Categories D, E, and F, horizontal cantilevers are designed for an upward force that results from an effective vertical acceleration of 1.2 times gravity. This is to provide some minimum strength in the upward direction and to account for possible dynamic amplification of vertical ground motions resulting from the vertical flexibility of the cantilever. The requirement is not applied to downward forces on cantilevers, for which the typical load combinations are used. C12.5 DIRECTION OF LOADING Seismic forces are delivered to a building through ground accelerations that may approach from any direction relative to the orthogonal directions of the building; therefore, seismic effects are expected to develop in both directions simultaneously. The standard requires structures to be designed for the most critical loading effects from seismic forces applied in any direction, and the procedures outlined in this section are deemed to satisfy that requirement. The orthogonal combination procedure combines the effects from 100 percent of the seismic load applied in one direction with 30 percent of the seismic load applied in the perpendicular direction. Combining effects for seismic loads in each direction and accidental torsion results in 16 load combinations as follows: Orthogonal load combinations where : QE = +/- QE_X+AT +/- 0.3QE_Y QE_Y = effect of Y-direction load at the center of mass (Section 12.8.4.2) QE = +/- QE_X-AT +/- 0.3QE_Y QE_X = effect of X-direction load at the center of mass (Section 12.8.4.2) QE = +/- QE_Y+AT +/- 0.3QE_X AT = accidental torsion computed in accordance with Section 12.8.4.2 QE = +/- QE_Y-AT +/- 0.3QE_X For horizontal structural elements such as beams and slabs, orthogonal effects may be minimal; however, for vertical elements of the seismic-force-resisting system that participate in both orthogonal directions, the design likely will be governed by these combinations. Orthogonal combinations should not be confused with modal combinations such as the square root of the sum of the squares (SRSS) or complete quadratic combination (CQC) technique. The maximum effect of seismic forces, QE, from orthogonal load combinations must be modified by the redundancy factor, ., or the overstrength factor, O0, and consider the effects of vertical seismic forces, EV, in accordance with Section 12.4, to obtain the seismic load effect, E. C12.6 ANALYSIS SELECTION PROCEDURE Table 12.6-1 applies only to buildings without seismic isolation (Chapter 17) or passive energy devices (Chapter 18). The procedures addressed in Table 12.6-1 are equivalent lateral force (ELF) analysis (Section 12.8), modal response spectrum (MRS) analysis (Section 12.9), linear response history (LRH) analysis, and nonlinear response history (NRH) analysis. Requirements for performing response history analysis are provided in Chapter 16. Nonlinear static (pushover) analysis is not addressed in the standard. The value of Ts (= SD1/SDS) depends on the site class because SDS and SD1 include such effects. Where ELF is not allowed, analysis must be performed using modal response spectrum or response history analysis. ELF is not allowed for buildings with the listed irregularities because it assumes a gradually varying distribution of mass and stiffness along the height and negligible torsional response. The 3.5Ts limit recognizes that higher modes are more significant in taller buildings (Lopez and Cruz, 1996; Chopra, 2007) such that the ELF method may underestimate the design base shear and may not predict correctly the vertical distribution of seismic forces. C12.7 MODELING CRITERIA C12.7.1 Foundation Modeling. Structural systems consist of three interacting components: the structural framing (girders, columns, walls, diaphragms), the foundation (footings, piles, caissons), and the supporting soil. The ground motion that a structure experiences, as well as the response to that ground motion, depends on the complex interaction between these components. Those aspects of ground motion that are affected by site characteristics are assumed to be independent of the structurefoundation system as these effects would occur in the free-field in the absence of the structure. Hence, site effects are considered separately (Sections 11.4.2 through 11.4.4 and Chapters 20 and 21). Given a site-specific ground motion or response spectrum, the dynamic response of the structure will depend on the foundation system and on the characteristics of the soil that support the system. The dependence of the response on the structure-foundation-soil system is referred to as soil-structure interaction. Such interactions will usually, but not always, result in a reduction of base shear. This reduction in shear is due to the flexibility of the foundation-soil system and an associated lengthening of the period of vibration of the structure. In addition, the soil system may provide an additional source of damping. However, that total displacement typically increases with soil-structure interaction. If the foundation is considered to be rigid, the computed base shears usually will be conservative, and it is for this reason that rigid foundation analysis is allowed. The designer may ignore soil-structure interaction or may consider it explicitly in accordance with Section 12.13.3 or implicitly in accordance with Chapter 19. C12.7.2 Effective Seismic Weight. During an earthquake, the structure accelerates laterally, and these accelerations of the structural mass produce inertial forces. These inertial forces, accumulated over the height of the structure, produce the design base shear. When a building vibrates during an earthquake, only that portion of the mass or weight that is physically tied to the structure needs to be considered as effective. Hence, live loads (e.g., loose furniture, loose equipment, and human occupants) need not be included. However, certain types of live loads such as storage loads may develop inertial forces, particularly where they are densely packed. Also considered as effective weight is all permanently attached equipment (e.g., air conditioners, elevator equipment, and mechanical systems), movable partitions (a minimum of 10 psf is required), and 20 percent of significant roof snow load. The full snow load need not be considered because maximum snow load and maximum earthquake load are unlikely to occur simultaneously and loose snow does not move with the roof. C12.7.3 Structural Modeling. The development of a mathematical model of a structure is always required because the story drifts and the design forces in the structure cannot be computed without such a model. In some cases, the mathematical model can be as simple as a free-body diagram as long that model can appropriately capture the strength and stiffness of the structure. The most realistic analytical model is three-dimensional, includes all sources of stiffness (and flexibility) of the structure and the soil-foundation system as well as P-delta effects, and allows for nonlinear inelastic behavior in all parts of the structurefoundation- soil system. Development of such an analytical model is very time consuming, and such analysis is rarely warranted for typical building designs performed in accordance with the standard. Instead of performing a nonlinear analysis, inelastic effects are accounted for indirectly in the linear analysis methods by means of the response modification factor, R, and the deflection amplification factor, Cd. Using modern software, it often is more difficult to decompose a structure into planar models than it is to develop a full three-dimensional model so three-dimensional models now are commonplace. Increased computational efficiency has reduced the motivation to model rigid diaphragms, allowing for easy and efficient modeling of diaphragm flexibility. Threedimensional models are required where the structure has torsional irregularities, out-of-plane offset irregularities, or nonparallel system irregularities. In general, the same three-dimensional model may be utilized for equivalent lateral force, modal response spectrum, and linear response history analysis. The response spectrum and linear response history models require a realistic modeling of structural mass, and the response history method also requires an explicit representation of inherent damping. Five percent critical damping is automatically included in the modal response spectrum approach. See Chapter 16 and the related commentary for additional information on linear and nonlinear response history analysis. It is well known that deformations in the panel zones of the beam-column joints of steel moment frames are a significant source of flexibility. Two different mechanical models for including such deformations are summarized in Charney and Marshall (2006). These methods apply to both elastic and inelastic systems. For elastic structures, centerline analysis provides reasonable, but not always conservative, estimates of frame flexibility. Fully rigid end zones should not be used, as this will always result in an overestimation of lateral stiffness in steel moment-resisting frames. Partially rigid end zones may be justified in certain cases such as where doubler plates are used to reinforce the panel zone. Including the effect of composite slabs on the stiffness of beams and girders may be warranted in some circumstances. Where composite behavior is included, due consideration should be paid to the reduction in effective composite stiffness for portions of the slab in tension (Schaffhausen and Wegmuller, 1977; Liew, et al., 2001) Figure C12.7-1 Undesired interaction effects. h H Expected plastic hinge capacity = M Expected column shear = 2M /H p p Actual column shear = 2M p /h Expected hinging region Unexpected hinging region For reinforced concrete buildings, it is important to address the effects of axial, flexural, and shear cracking in modeling the effective stiffness of the structural components. Determining appropriate effective stiffness of the structural components should take into consideration the anticipated demands on the components, their geometry, and the complexity of the model. Recommendations for computing cracked section properties may be found in Paulay and Priestley (1992) and similar texts. C12.7.4 Interaction Effects. The interaction requirements are intended to prevent unexpected failures in members of moment-resisting frames. Figure C12.7-1 illustrates a typical situation where masonry infill is used, and this masonry is fitted tightly against reinforced concrete columns. Since the masonry is much stiffer than the columns, column hinges form at the top of column and at the top of the masonry rather than at the top and bottom of the column. If the column flexural capacity is Mp, the shear in the columns increases by the factor H/h, and this may cause an unexpected nonductile shear failure in the columns. Many building collapses have been attributed to this effect. Figure C12.7-1 Undesired interaction effects. C12.8 EQUIVALENT LATERAL FORCE PROCEDURE The equivalent lateral force (ELF) procedure provides a simple way to incorporate the effects of inelastic dynamic response into a linear static analysis. This procedure is useful in preliminary design of all structures and is allowed for final design of the vast majority of structures. The procedure is valid only for structures without significant discontinuities in mass and stiffness along the height, where the dominant response to ground motions is in the horizontal direction without significant torsion. The ELF procedure has three basic steps: 1. Determine the seismic base shear, 2. Distribute the shear vertically along the height of the structure, and 3. Distribute the shear horizontally across the width and breadth of the structure. Each of these steps is based on a number of simplifying assumptions. A broader understanding of these assumptions may be obtained from any structural dynamics textbook that emphasizes seismic applications. C12.8.1 Seismic Base Shear C12.8.1.1 Calculation of Seismic Response Coefficient. Equation 12.8-1 simply expresses the base shear as the product of the effective seismic weight, W, and a response coefficient, Cs. The response coefficient is a spectral pseudoacceleration, in g units, which has been modified by R and I to account for inelastic behavior and to provide for improved performance for high occupancy or essential structures. There are five equations for determining the response coefficient Cs; the first three are plotted in Figure C12.8-1. Equation 12.8-2, representing the constant acceleration part of the spectrum, controls where 0.0 < T < Ts. As shown in Table C12.6-1 (which provides values of 3.5Ts), Ts is a function of seismicity and site. It may be as low as 0.2 seconds for low hazard regions on Site Class B or as high as 0.9 seconds in high hazard regions on Site Class E. Figure C12.8-1 Seismic response coefficient versus period. T0 TS TL Seismic R esponse Coefficient, Cs Period, T Constant acceleration [Eq. 12.8-2] Constant velocity [Eq. 12.8-3] Constant displacement [Eq. 12.8-4] Transition to peak ground acceleration [not used for ELF] The true pseudoacceleration response spectrum transitions to the peak ground acceleration as the period approaches zero. This transition is not used in the ELF method. One reason is that simple reduction of the response spectrum by (1/R) in the very short period region would exaggerate inelastic effects. Figure C12.8-1 Seismic response coefficient versus period. Equation 12.8-3, representing the constant velocity part of the spectrum, controls where Ts < T < TL. In this region, the seismic response coefficient is inversely proportional to period, and the pseudovelocity (pseudoacceleration divided by circular frequency, lower case omega, is constant. TL, the long-period transition period, is provided in Figures 22-15 through 22-20. TL ranges from 4 seconds in in the northcentral conterminous states and western Hawaii to 16 seconds in the Pacific Northwest and in western Alaska. Equation 12.8-4, representing the constant displacement part of the spectrum, controls where T > TL. Given the current mapped values of TL, this equation only affects tall and flexible structures. Equation 12.8-5 is the minimum base shear and provides a (working stress) strength of approximately 3 percent of the weight of the structure (Seismology Committee, Structural Engineers Association of California, 1996). This minimum base shear was originally enacted in 1933 by the state of California’s Riley Act. Equation 12.8-6 applies to sites near major active faults (as reflected by values of S1) where pulse effects can increase longperiod demands. C12.8.1.2 Soil-Structure Interaction Reduction. Soil-structure interaction, which can influence significantly the dynamic response of structures to earthquakes, is addressed in Chapter 19. C12.8.1.3 Maximum Ss Value in Determination of Cs. The maximum value of Ss was created as hazard maps were revised in 1997. The cap on Ss reflects engineering judgment about performance of code-complying buildings in past earthquakes so the height, period, and regularity conditions required for use of the limit are very important qualifiers. C12.8.2 Period Determination. The fundamental period of the structure, T, is used to determine the design base shear as well as the exponent k that establishes the distribution of the shear along the height of the structure. Equation 12.8-7 is an Figure C12.8-2 Variation of fundamental period with building height. 0 1 2 3 4 5 6 7 0 100 200 300 400 500 600 Fundamental period, T (s) Building height, hn (ft) measured values [mean minus one standard deviation] [mean] 0.028 0.80 a n T = h 0.035 0.80 a n T = h empirical relationship determined through statistical analysis of the measured response of buildings in California. Figure C12.8-2 illustrates such data for various structures with steel moment resisting frames. Since the empirical expression is based on the lower bound of the data, it produces a lower bound for the period of a building of given height. This lower bound period, used in Equations 12.8-3 and 12.8-4, provides a conservative estimate of base shear. The fundamental period determined from a rational analysis may be used in design unless it exceeds the approximate period times the coefficient provided in Table 12.8-1. This period limit prevents the use of unusually low ELF base shear for design of buildings (or computational models) that are overly flexible. The coefficients in the table have two effects. First, the conservatism of lower bound empirical formulas for Ta is removed. Second, the period is increased in regions of lower seismicity as buildings in such areas generally are more flexible (and, hence, have longer periods) than buildings in regions of higher seismicity. Figure C12.8-2 Variation of fundamental period with building height. C12.8.3 Vertical Distribution of Seismic Force. Equation 12.8-12 is based on the simplified first mode shape shown in Figure C12.8-3. In the figure, Fx is the inertial force at level x, which is simply the total acceleration at level x times the mass at level x. The base shear is the sum of these inertial forces, and Equation 12.8 simply gives the ratio of the force at level x to the total base shear. The deformed shape of the structure of Figure C12.8-3 is a function of the exponent k, which is related to the fundamental period of vibration of the structure. The variation of k with T is illustrated in Figure C12.8-4. The exponent k is intended to approximate the effect of higher modes, which are generally more dominant in structures with a longer fundamental period of vibration. Lopez and Cruz (1996) discuss the factors that influence higher modes of response. Although the actual first mode shape for a structure is also a function of the type of seismic-force-resisting system, that effect is not reflected in these equations. The horizontal forces computed using Equation 12.8-12 do not reflect the actual inertial forces imparted on a structure at any particular time. Instead, they are intended to provide design story shears that are consistent with enveloped results from more accurate analysis (Chopra and Newmark, 1980). Building height, hn (ft) Figure C12.8-3 Basis of Equation 12.8-12. hx k x wx h Figure C12.8-3 Basis of Equation 12.8-12. 2 2 1 k x x x n k i b i i k x x x vx n b k i i F h w g V h w g C F w h V wh . a . a = = = = = S S Figure C12.8-4 Variation of exponent k with period T. k k = 0.75 + 0.5T T (seconds) 1.0 2.0 0.5 2.5 Figure C12.8-3 Basis of Equation 12.8-12. C12.8.4 Horizontal Distribution of Forces. Within the context of an elastic ELF analysis, the distribution of lateral forces to various seismic-force-resisting elements depends on the type, geometric arrangement, and vertical extents of the resisting elements and on the shape and flexibility of the floor diaphragms. Because seismic-force-resisting elements are expected to respond inelastically to design ground motions, the distribution of forces to the various elements also depends on the strength of the elements and their sequence of yielding. Clearly, such effects cannot be captured accurately by a linear elastic static analysis (Paulay, 1997). Nonlinear dynamic analysis is too cumbersome to be applied to the design of most buildings so other approximate methods are used. Figure C12.8-4 Variation of exponent k with period T. Of particular concern is the torsional response of the structure during the earthquake. This response has been observed in structures that are designed to be nearly symmetric in plan and layout of seismic-force-resisting systems (De La Liera and Chopra, 1994). This torsional response is due to a variety of “accidental” eccentricities that exist due to uncertainties in quantifying the mass and stiffness distribution of the structure, as well as torsional components of ground motion that are not included explicitly in code-based designs (Newmark and Rosenbleuth, 1971). C12.8.4.1 Inherent Torsion. When lateral forces in a particular direction are applied statically at each story of a building with rigid diaphragms, torsional displacement (twisting about the vertical axis) occurs if the centers of stiffness and mass of each story are not perfectly coincident in plan. When three-dimensional analysis is used, this inherent torsion is included automatically. When planar analysis is used, the centers of mass and rigidity for each story must be determined explicitly. Unfortunately, it is difficult to determine the center of rigidity for a multistory building to compute the inherent torsion; the center of rigidity for a particular story depends on the configuration of the seismic-force-resisting elements above and below that story and may be load dependent (Chopra and Goel, 1991). For buildings with fully flexible diaphragms (as defined in Section 12.3), vertical elements are assumed to resist inertial forces from the mass that is tributary to the elements, but with no explicitly computed torsion. No diaphragm is perfectly flexible, so some torsional forces always develop even when they are ignored. Figure C12.8-5 Amplification factor for symmetric rectangular buildings. L/B = 4 L/B = 1 cap floor 0 1 2 3 0.1 0.2 0.3 0.4 Torsional amplification factor, Ax Dimensional coefficient, a curves for L/B = 1, 1. , 2, 4 L B aL aB V C12.8.4.2 Accidental Torsion. Even for perfectly symmetric buildings, the true locations of the centers of mass and rigidity are uncertain. As discussed in Section C12.8.4, other effects also may produce torsion. The requirement to consider accidental torsion is intended to address this concern. Accidental and inherent torsions result in forces that must be combined with those obtained from the application of the lateral story forces; all components must be designed for the maximum effects determined considering positive accidental torsion, negative accidental torsion, and no accidental torsion. C12.8.4.3 Amplification of Accidental Torsion. Equation 12.8-14 was developed by the SEAOC “seismology committee to encourage buildings with good torsional stiffness” (Structural Engineers Association of California, 1999). In calculating the torsional amplification factor, Ax, the applied loads include inherent and accidental torsion, but with no further amplification; the calculation is not iterative. Figure C12.8-5 illustrates the effect of Equation 12.8-14 for a symmetric rectangular building with various aspect ratios (L/B) where the seismic-force-resisting elements are positioned at a variable distance (defined by a) from the center of mass in each direction. Each element is assumed to have the same stiffness. The structure is loaded parallel to the short direction with an eccentricity of 0.05L. Figure C12.8-5 Amplification factor for symmetric rectangular buildings. For a equal to 0.5, these elements are at the perimeter of the buildings, and for a equal to 0.0, they are at the center (providing no torsional resistance). For a square building (L/B = 1.00), the torsional amplification factor is greater than 1.0 where a is less than 0.25 and increases to the maximum of 3.0 where a is equal to 0.11. For a rectangular building with L/B equal to 4.00, the amplification factor is greater than 1.0 where a is less than 0.34 and increases to 3.0 where a is equal to 0.15. For the range of aspect ratios investigated, Ax is equal to 1.0 where a is greater than 0.34 and Ax reaches its maximum value of 3.0 where (a < 0.11 to 0.15). C12.8.6 Story Drift Determination. Equation 12.8-15 is used to estimate inelastic deflections, which are then used to calculate design story drifts. These story drifts must be less than the allowable story drifts of Table 12.12-1. For buildings without torsional irregularity, computations are performed using deflections at the centers of mass of adjacent stories. For Seismic Design Category C, D, E, or F structures that are torsionally irregular, Section 12.12.1 requires that drifts be computed along the edges of the structure. Figure C12.8-6 Displacements used to compute drift. Force, V Displacement, Elastic response Actual inelastic response Idealized inelastic response VE V=V /R Analysis domain E d d E d The term Cd in Equation 12.8-15 amplifies the displacements from elastic analysis at design level forces, which are reduced by R. Figure C12.8-6 illustrates the relationships between elastic response; response to reduced design-level forces; and the expected inelastic response. If the structure remained elastic during an earthquake, the force would be VE, and the corresponding displacement would be d E. Note that VE does not include the reduction factor, R, which accounts primarily for ductility and overstrength. According to the equal displacement “rule” of seismic design, the maximum displacement of an inelastic system is approximately equal to that of an elastic system with the same initial stiffness. This condition has been observed for structures idealized with bilinear inelastic response and a fundamental period greater than Ts. For shorter period structures, peak displacement of an inelastic system tends to exceed that of the corresponding elastic system. Since the forces used for design include the response modification coefficient, R, the resulting displacements are too small and must be amplified. Figure C12.8-6 Displacements used to compute drift. This analysis domain is shown in Figure C12.8-6. Because of overstrength and associated stiffness increases, the actual inelastic response differs from the idealized inelastic response; the actual displacement of the system may be less than R times d. The standard accounts for this difference by multiplying the fictitious (design-level) elastic displacements d by the factor Cd, which is usually less than R. The design forces used to compute d xe include the importance factor, I, so Equation 12.8-15 includes I in the denominator. This is appropriate since the allowable story drifts (except for masonry shear wall structures) in Table 12.12-1 are more stringent for higher occupancy categories. C12.8.6.1 Minimum Base Shear for Computing Drift. Except for period limits (as described in Section C12.8.6.2), all of the requirements of Section 12.8 (including minimum base shears and force distributions) must be satisfied where computing drift for ELF analysis. C12.8.6.2 Period for Computing Drift. Where the response spectrum of Section 11.4.5 or the corresponding equations of Section 12.8.1 are used and the structural period is less than TL, displacements increase with increasing period (even though forces may decrease). Section 12.8.2 applies a period limit so that design forces are not too low, but if the lateral forces used to compute drifts are inconsistent with the forces corresponding to the computed period, displacements will be overestimated. Therefore, the standard allows the determination of drift using forces that are consistent with the computed period of vibration of the structure. Computed periods greater than CuTa are common, particularly for moment frames. In such cases the seismic design forces used to proportion strength may produce displacements that violate drift limits, whereas displacements based on the computed period will satisfy drift limits. Equation 0 G 0y 0y K P K V h d . = = Equation (1 ) 1y 0y V =V -. Equation 0 1 1 d d . = - Figure C12.8-7 P-delta effect on a simple structure. Displacement, Force, V d y (= d 0y = d 1y) V0 y V1 y Slope = K0 Slope = K 1 = K 0 + K G Excluding P-delta Including P-delta Slope = KG d d 0 d 1 The more flexible the structure, the more likely it is that P-delta effects will ultimately control the design. Computed periods that are significantly greater than (perhaps more than 1.5 times) CuTa may indicate a modeling error. C12.8.7 P-delta Effects. P-delta effects influence both the stiffness and strength of structures. Figure C12.8-7 shows idealized static force-displacement responses for a simple, one-story structure (such as a cantilevered column). The stiffness and strength of the structure without considering P-delta effects (condition 0) are represented by K0 and V0. When P-delta effects are considered (condition 1), the related quantities are K1 and V1. Since the two model conditions are for the same structure, inherent capacity of the structure is the same in either condition, the yield displacement is the same (d 0y = d 1y = d y). The geometric stiffness of the structure, KG, is equal to P/h, where P is the total gravity load and h is the story height. KG is negative where gravity loads cause compression in the story. The stability coefficient, ., is defined as the absolute value of the geometric stiffness divided by the elastic stiffness. From Figure C12.8-7, K0 = V0y / d 0y. Hence, C12.8-1 Given the above, and the geometric relationships shown in Figure C12.8-7, it can be shown that the force producing yield in condition 1 (with P-delta effects) is C12.8-2 and that for an applied force, V, less than or equal to V1y C12.8-3 As . approaches 1.0, d 1 approaches infinity and V1 approaches zero, defining a state of static instability. Figure C12.8-7 P-delta effect on a simple structure. The intent of Section 12.8.7 is to determine whether P-delta effects are significant, and if so, to modify the strength and stiffness of the structure to account for such effects. Also, maximum permitted values of . are established. Equation 12.8-16 is used to determine the stability coefficient of each story of a structure. Where the stability coefficient exceeds 0.1, P-delta effects must be considered using one of two approaches. Displacements and member forces are either multiplied by 1/(1-. ) to reflect the conditions shown in Figure C12.8-7 in accordance with the equal displacement rule or determined by rational analysis. Two types of rational analysis are envisioned. First, a nonlinear static (pushover) analysis could be performed to show that the post-yield slope of the pushover curve is continuously positive up to the target displacement. Second, a nonlinear dynamic response history analysis could be repeated with and without P-delta effects to determine if the behavior including P-delta meets all performance criteria. Although the P-delta procedures in the standard reflect the simple static idealization shown in Figure C12.8-7, the real issue is one of dynamic stability. For that reason, nonlinear response history analysis is appealing. Such analysis should reflect variability of ground motions and system properties, including initial stiffness, strain hardening stiffness, initial strength, hysteretic behavior, and magnitude of gravity load. Unfortunately, the dynamic response of structures is highly sensitive to such parameters, causing considerable dispersion to appear in the results (Vamvatsikos, 2002). This dispersion, which increases dramatically with stability coefficient ., is due primarily to the incrementally increasing residual deformations (ratcheting) that occur during the response. Residual deformations may be controlled by increasing either the initial strength or the secondary stiffness. See Gupta and Krawinkler (2000) for additional information. Equation 12.8-17 establishes the maximum stability coefficient permitted. The intent of this requirement is to protect structures from the possibility of stability failures triggered by post-earthquake residual deformation. C12.9 MODAL RESPONSE SPECTRUM ANALYSIS In the modal response spectrum analysis method, the structure is decomposed into a number of single-degree-of-freedom systems, each having its own mode shape and natural period of vibration. The number of modes available is equal to the number of mass degrees of freedom of the structure, so the number of modes can be reduced by eliminating mass degrees of freedom. For example, rigid diaphragm constraints may be used to reduce the number of mass degrees of freedom to one per story for planar models, and to three per story (two translations and rotation about the vertical axis) for three-dimensional structures. However, where the vertical elements of the seismic-force-resisting system have significant differences in lateral stiffness, rigid diaphragm models should be used with caution as relatively small in-plane diaphragm deformations can have a significant effect on the distribution of forces. For a given direction of loading, the displacement in each mode is determined from the corresponding spectral acceleration, modal participation, and mode shape. Because the sign (positive or negative) and the time of occurrence of the maximum acceleration are lost in creating a response spectrum, there is no way to recombine modal responses exactly. However, statistical combination of modal responses produces reasonably accurate estimates of displacements and component forces. The loss of signs for computed quantities leads to problems in interpreting force results where seismic effects are combined with gravity effects, produces forces that are not in equilibrium, and makes it impossible to plot deflected shapes of the structure. C12.9.1 Number of Modes. The key motivation to perform modal response spectrum analysis is to determine how the actual distribution of mass and stiffness of a structure affects the elastic displacements and component forces. Where at least 90 percent of the model mass participates in the response, the distribution of forces and displacements is sufficient for design. The scaling required by Section 12.9.4 controls the overall magnitude of design values so that incomplete mass participation does not produce unconservative results. The number of modes required to achieve 90 percent modal mass participation is usually a small fraction of the total number of modes. See Lopez and Cruz (1996) for further discussion of the number of modes to use for modal response spectrum analysis. C12.9.2 Modal Response Parameters. The design response spectrum (whether the general spectrum from Section 11.4.5 or a site-specific spectrum determined in accordance with Section 21.2) is representative of linear elastic structures. Division of the spectral ordinates by R accounts for inelastic behavior, and multiplication of spectral ordinates by I provides the additional strength needed to improve the performance of important structures. The displacements that are computed using the response spectrum that has been modified by R and I (for strength) must be amplified by Cd and reduced by I to produce the expected inelastic displacements. (See Section C12.8.6.) C12.9.3 Combined Response Parameters. Most computer programs provide for either the SRSS or the CQC method (Wilson, et al., 1981) of modal combination. The two methods are identical where applied to planar structures, or where zero damping is specified for the computation of the cross-modal coefficients in the CQC method. The modal damping specified in each mode for the CQC method should be equal to the damping level that was used in the development of the response spectrum. For the spectrum in Section 11.4.5, the damping ratio is 0.05. The SRSS or CQC method is applied to loading in one direction at a time. Where Section 12.5 requires explicit consideration of orthogonal loading effects, the results from one direction of loading may be added to 30 percent of the results from loading in an orthogonal direction. Wilson (2000) suggests that a more accurate approach is to use the SRSS method to combine 100 percent of the results from each of two orthogonal directions where the individual directional results have been combined by SRSS or CQC, as appropriate. C12.9.4 Scaling Design Values of Combined Response. The modal base shear, Vt, may be less than the ELF base shear, V, because: (a) the calculated fundamental period may be longer than that used in computing V, (b) the response is not characterized by a single mode, and (c) the ELF base shear assumes 100 percent mass participation in the first mode, which is always an overestimate. The scaling required by Section 12.9.4 provides, in effect, a minimum base shear for design. This minimum base shear is provided because the computed period of vibration may be the result of an overly flexible (incorrect) analytical model. The possible 15 percent reduction in design base shear may be considered as an incentive for using a modal response spectrum analysis in lieu of the equivalent lateral force procedure. Displacements from the modal response spectrum are not scaled because the use of an overly flexible model will result in conservative estimates of displacement that need not be further scaled. C12.9.5 Horizontal Shear Distribution. Accidental torsion must be included in the analysis as specified in Section 12.8.7. For modal analysis there are two basic approaches to include accidental torsion. The first approach is to perform static analyses with accidental torsions applied at each level of the structure, and then add these results to those obtained from the modal response spectrum analysis. Where this approach is used, torsional amplification in accordance with Section 12.8.4.3 is required. The second approach, which applies only to three-dimensional analysis, is to offset the centers of mass of each story 5 percent in each direction, thus requiring four separate models. The advantage of this method is that the effects of direct loading and accidental torsion are combined automatically. A practical disadvantage is the increased bookkeeping for multiple analyses. Where this approach is used, further amplification of accidental torsion is not required because repositioning the center of mass in a dynamic analysis changes the natural mode shapes and frequencies, producing torsions larger than the static accidental torsion. C12.9.6 P-delta Effects. The requirements of Section 12.8.7, including the stability coefficient limit, . max, apply to modal response spectrum analysis. Amplification of displacements and member forces as a result of P-delta effects may be accomplished through use of the geometric stiffness. For the purpose of dynamic analysis, the linearized geometric stiffness, which includes the story-wise P- . effect, is usually sufficient. Using the consistent geometric stiffness (P-d effect), which is associated with the deflected shape of the individual elements of the structure, slightly improves accuracy. Including P-delta effects directly in dynamic analysis lengthens of the periods of vibration of each mode of response and increases lateral displacements. C12.10 DIAPHRAGMS, CHORDS, AND COLLECTORS C12.10.1 Diaphragm Design. Diaphragms are generally treated as horizontal deep beams or trusses that distribute lateral forces to the vertical elements of the seismic-force-resisting system. As deep beams, diaphragms must be designed to resist the resultant shear and bending stresses. Diaphragms are commonly compared to girders, with the roof or floor deck analogous to the girder web in resisting shear, and the boundary elements (chords) analogous to the flanges of the girder in resisting flexural tension and compression. As in girder design, the chord members (flanges) must be sufficiently connected to the body of the diaphragm (web) to prevent separation and to force the diaphragm to work as single unit. Diaphragms may be considered flexible, semi-rigid, or rigid. The flexibility or rigidity of the diaphragm determines how lateral forces will be distributed to the vertical elements of the seismic-force-resisting system. See Section C12.3.1. Once the distribution of lateral forces is determined, shear and moment diagrams are used to compute the diaphragm shear and chord forces. Where diaphragms are not flexible, inherent and accidental torsion must be considered in accordance with Section 12.8.4. Diaphragm openings may require additional localized reinforcement (sub-chords and collectors) to resist the subdiaphragm chord forces above and below the opening and to collect shear forces where the diaphragm depth is reduced. (See Figure C12.10-1.) Collectors on each side of the opening drag shear into the subdiaphragms above and below the opening. The subchord and collector reinforcement must extend far enough into the adjacent diaphragm to develop the axial force through shear transfer. The required development length is determined by dividing the axial force in the sub-chord by the shear capacity (in force/unit length) of the main diaphragm. Chord reinforcement at reentrant corners must extend far enough into the main diaphragm to develop the chord force through shear transfer. (See Figure C12.10-2.) Continuity of the chord members also must be considered where the depth of the diaphragm is not constant. Figure C12.10-1 Diaphragm components. Main diaphragm chords Subdiaphragm Collector elements Shear wall Shear wall Subdiaphragm Direction of loading Sub-chords Figure C12.10-2 Diaphragm with a re-entrant corner. Main diaphragm chords Shear wall Shear wall Direction of loading Chord force development length In wood and metal deck diaphragm design, framing members are often used as continuity elements, serving as sub-chords and collector elements at discontinuities. These continuity members also are often used to transfer wall out-of-plane forces to the main diaphragm, where the diaphragm itself does not have the capacity to resist the anchorage force directly. For additional discussion, see Section C12.11.2.2.3. Figure C12.10-1 Diaphragm components. Figure C12.10-2 Diaphragm with a re-entrant corner. C12.10.1.1 Diaphragm Design Forces. Diaphragms must be designed to resist inertial forces, as specified in Equation 12.10-1, and to transfer design seismic forces due to horizontal offsets or changes in stiffness of the vertical resisting elements. Inertial forces are those seismic forces that originate at the specified diaphragm level, while the transfer forces originate above the specified diaphragm level. The redundancy factor, ., used for design of the seismic-force-resisting elements also applies to diaphragm transfer forces, thus completing the load path. C12.10.2.1 Collector Elements Requiring Load Combinations with Overstrength Factor for Seismic Design Categories C through F. The overstrength requirement of this section is intended to keep inelastic behavior in the ductile elements of the seismic-force-resisting system (consistent with the R factor) rather than in collector elements. C12.11 STRUCTURAL WALLS AND THEIR ANCHORAGE As discussed in Section C11.7, structural integrity is important not only in earthquake-resistant design but also in resisting high winds, floods, explosion, progressive failure, and even such ordinary hazards as foundation settlement. The detailed requirements of this section address wall-to-diaphragm integrity. C12.11.1 Design for Out-of-Plane Forces. Because they are often subjected to local deformations caused by material shrinkage, temperature changes, and foundation movements, wall connections require some degree of ductility in order to accommodate slight movements while providing the required strength. Although nonstructural walls are not subject to this requirement, they must be designed in accordance with Chapter 13. C12.11.2 Anchorage of Concrete or Masonry Structural Walls. One major hazard in past earthquakes is the separation of heavy masonry or concrete walls from floors or roofs. The forces defined in this section apply only to the anchorage or connection of the wall to the structure, and not to overall wall design. The anchorage force should be considered both for tension (out-of-plane) and sliding (in-plane) directions. Where the lateral spacing of connections used to resist the wall anchorage force are spaced further apart than 4 feet (1219 mm) as measured along the length of the wall, the section of wall that spans between the anchors must be designed to resist the local out-of-plane bending caused by this force. C12.11.2.1 Anchorage of Concrete or Masonry Structural Walls to Flexible Diaphragms. Diaphragm flexibility can amplify out-of-plane accelerations so the wall anchorage forces in this condition are twice those defined in Section 12.11.1. C12.11.2.2 Additional Requirements for Diaphragms in Structures Assigned to Seismic Design Categories C through F. C12.11.2.2.1 Transfer of Anchorage Forces into Diaphragm. This requirement, which aims to prevent the diaphragm from tearing apart during strong shaking by requiring transfer of anchorage forces across the complete depth of the diaphragm, was prompted by failures of connections between tilt up concrete walls and wood panelized roof systems in the 1971 San Fernando earthquake. An exception is provided for modestly proportioned diaphragms of light-frame construction, which have not performed poorly. Depending upon diaphragm shape and member spacing, numerous suitable combinations of subdiaphragms and continuous tie elements and smaller sub-subdiaphragms connecting to larger subdiaphragm and continuous tie elements are possible. The configuration of each subdiaphragm (or sub-subdiaphragm) provided must comply with the simple 2.5-to-1 length-towidth ratio, and the continuous ties must have adequate member and connection strength to carry the accumulated wall anchorage forces. C12.11.2.2.2 Steel Elements of Structural Wall Anchorage System. A multiplier of 1.4 has been specified for strength design of steel elements in order to obtain a fracture strength of almost 2 times the specified design force (where ft is 0.75 for tensile rupture). C12.11.2.2.3 Wood Diaphragms. Material standards for wood structural panel diaphragms permit the sheathing to resist shear forces only; use to resist direct tension or compression forces is not permitted. Therefore, seismic anchorage forces from walls must be transferred into framing members (such as beams, purlins, or subpurlins) using suitable straps or anchors. For wood diaphragms, it is common to use local framing and sheathing elements as subdiaphragms to transfer the uniform lateral wall forces into more concentrated lines of drag or continuity framing that carry the forces across the diaphragm and hold the building together. Figure C12.11-1 shows a schematic plan of typical roof framing using subdiaphragms. Fasteners to wood framing are intended to transfer shear forces only along the wood framing; any forces acting transverse to the framing tend to induce splitting (due to cross-grain tension). Fasteners into wood ledgers attached to concrete or masonry walls are designed to resist shear forces only; separate straps or anchors generally are provided to transfer out-of-plane wall forces into perpendicular framing members. C12.11.2.2.4 Metal Deck Diaphragms. In addition to transferring shear forces, metal deck diaphragms often can resist direct axial forces in at least one direction. However, corrugated metal decks cannot transfer axial forces in the direction perpendicular to the corrugations and are prone to buckling if the unbraced length of the deck as a compression element is large. To manage diaphragm forces perpendicular to the deck corrugations, it is common that metal decks are supported at 8- to 10-foot intervals by joists that are connected to walls in a manner suitable to resist the full wall anchorage design force and to carry that force across the diaphragm. In the direction parallel to the deck corrugations, subdiaphragm systems are considered near the walls; if the compression forces in the deck become large relative to the joist spacing, small compression reinforcing elements are provided to transfer the forces into the subdiaphragms. C12.11.2.2.6 Eccentrically Loaded Anchorage System. Wall anchors often are loaded eccentrically, either because the anchorage mechanism allows eccentricity, or because of anchor bolt or strap misalignment. This eccentricity reduces the anchorage connection capacity and hence must be considered explicitly in design of the anchorage. Figure C12.11-2 shows a one-sided roof-to-wall anchor that is subjected to severe eccentricity due to a misplaced anchor rod. If the detail were designed as a concentric two-sided connection, this condition would be easier to correct. C12.11.2.2.7 Walls with Pilasters. The anchorage force at pilasters must be calculated considering two-way bending in wall panels. It is customary to anchor the walls to the diaphragms assuming one-way bending and simple supports at the top and bottom of the wall. However, where pilasters are present in the walls, their stiffening effect must be taken into account. Each panel between pilasters is supported on four sides. The reaction at the pilaster top is the result of two-way action of the Figure C12.11-1 Typical subdiaphragm framing. Girder line (typically also continuity ties) Sub-diaphragm chords Sub-diaphragm Main diaphragm #1 Main diaphragm #1 chords Purlins (typical) Sub-diaphragm chords Main diaphragm #2 Typical sub-diaphragm for out-of-plane forces Opening Opening Figure C12.11-2 Plan view of wall anchor with misplaced anchor rod. Roof joist Shim added due to misplaced anchor rod Hold-down anchor Cast-in-place anchor rod Tilt-up wall panel Alternate solution to one sided connection: use two-sided connection panel and is applied directly to the beam or girder anchorage at the top of the pilaster. The anchor load at the pilaster generally is larger than the typical uniformly distributed anchor load between pilasters. Figure C12.11-3 shows the tributary area typically used to determine the anchorage force for a pilaster. Anchor points adjacent to the pilaster must be designed for the full tributary loading, conservatively ignoring the effect of the adjacent pilaster. Figure C12.11-1 Typical subdiaphragm framing. Figure C12.11-2 Plan view of wall anchor with misplaced anchor rod. C12.12 DRIFT AND DEFORMATION As used in the standard, deflection is the absolute lateral displacement of any point in a structure relative to its base, and story drift is the difference in deflection across a story (i.e., the deflection of a floor relative to that of the floor below). The drifts and deflections are checked for the design earthquake ground motion, which is two-thirds of the maximum considered earthquake (MCE) ground motion. Figure 12.11-3 Tributary area used to determine anchorage force at pilaster. Tributary area of wall on pilaster for pilaster anchorage design Wall yield line 45° Top of parapet Roof line Pilaster anchorage There are many reasons to control drift; the most significant are to address the structural performance concerns of member inelastic strain and system stability and to limit damage of nonstructural components, which can be life-threatening. Drifts provide a direct but imprecise measure of member strain and structural stability. Under small lateral deformations, secondary stresses due to the P-delta effect are normally within tolerable limits. (See Section C12.8.7.) The drift limits provide indirect control of structural performance. Figure 12.11-3 Tributary area used to determine anchorage force at pilaster. Buildings subjected to earthquakes need drift control to restrict damage of partitions, shaft and stair enclosures, glass, and other fragile nonstructural elements. The drift limits have been established without regard to economic considerations such as a comparison of present worth of future repairs with additional structural costs to limit drift. These are matters for building owners and designers to address. The drift limits of Table 12.12-1 reflect consensus judgment taking into account life safety and damage control objectives described above. Since the displacements induced in a structure include inelastic effects, structural damage in the designlevel earthquake is likely. This may be seen from the seismic drift limits stated in Table 12.12-1. For ordinary structures (Occupancy Category I or II), the drift limit is 0.02hsx, which is about ten times the drift ordinarily allowed under wind loads. If deformations well in excess of the seismic drift limits were to occur repeatedly, structural components could lose so much stiffness or strength that they compromise the safety and stability of the structure. To provide better performance for Occupancy Category IV essential facilities, their drift limits generally are more stringent than those for Occupancy Categories II and III. However, those limits are still greater than the damage thresholds for most nonstructural components. Therefore, while the performance of Occupancy Category IV buildings should be better than that of lower Occupancy Category buildings, there still can be considerable damage in the design earthquake. The drift limits for low-rise structures are relaxed somewhat, provided that the interior walls, partitions, ceilings, and exterior wall systems have been designed to accommodate story drifts. The type of steel building envisioned by the exception to the table would be similar to a prefabricated steel structure with metal skin. The limits set forth in Table 12.12-1 are for story drifts and apply to each and every story. For some structures, satisfying strength requirements may produce a system with adequate drift control. However, the design of moment-resisting frames and of tall, narrow shear walls or braced frames often is governed by drift considerations. Where design spectral response accelerations are large, seismic drift considerations are expected to control the design of midrise buildings. Where design spectral response accelerations are small or the building is very tall, design for wind generally will control. C12.12.3 Building Separation. The intent of this section is to address separations (also called seismic joints) between adjacent structures or portions of the same structure (with or without frangible closures) for the purpose of permitting independent response to earthquake ground motion. For irregular structures that cannot be expected to act reliably as a unit, seismic joints should be used to produce separate units whose independent response to earthquake ground motion can be predicted. The standard does not give a precise formulation for the separations, but it does require that the distance be “sufficient to avoid damaging contact under total deflection.” It is recommended that the distance be no less than the square root of the sum of the squares of the lateral deflections, which represent the anticipated maximum inelastic deformations including torsion, of the two units assumed to deflect toward each other (thus increasing with height). If the effects of impact can be shown not to be detrimental, these distances can be reduced. For very rigid shear wall structures with rigid diaphragms whose lateral deflections cannot be reasonably estimated, it is suggested that older code requirements for structural separations of at least 1 inch (25 mm) plus 1/2 inch (13 mm) for each 10 feet (3 m) of height above 20 feet (6 m) be followed. C12.12.4 Deformation Compatibility For Seismic Design Categories D Through F. The purpose of this section is to require that the seismic-force-resisting system provide adequate deformation control to protect elements of the structure that are not part of the seismic-force-resisting system. In regions of high seismicity, many designers apply ductile detailing requirements to elements that are intended to resist seismic forces but neglect such practices in nonstructural elements or elements intended to resist only gravity forces. Even where elements of the structure are not intended to resist seismic forces and are not detailed for such resistance, they can participate in the response and suffer severe damage as a result. In the 1994 Northridge earthquake, such participation was a cause of several failures. A preliminary reconnaissance report of that earthquake (EERI, 1994) states: Of much significance is the observation that six of the seven partial collapses (in modern precast concrete parking structures) seem to have been precipitated by damage to the gravity load system. Possibly, the combination of large lateral deformation and vertical load caused crushing in poorly confined columns that were not detailed to be part of the lateral load resisting system. . . . Punching shear failures were observed in some structures at slab-to-column connections such as at the Four Seasons building in Sherman Oaks. The primary lateral load resisting system was a perimeter ductile frame that performed quite well. However, the interior slab-column system was incapable of undergoing the same lateral deflections and experienced punching failures. This section addresses such concerns. Rather than relying on designers to assume appropriate levels of stiffness, this section explicitly requires that the stiffening effects of adjoining rigid structural and nonstructural elements be considered and that a rational value of member and restraint stiffness be used for the design of components that are not part of the seismic-forceresisting system. This section also includes a requirement to address shears that can be induced in structural components that are not part of the seismic-force-resisting system, since sudden shear failures have been catastrophic in past earthquakes. The exception in Section 12.12.4 is intended to encourage the use of intermediate or special detailing in beams and columns that are not part of the seismic-force-resisting system. In return for better detailing, such beams and columns are permitted to be designed to resist moments and shears from unamplified deflections. This reflects observations and experimental evidence that well-detailed components can accommodate large drifts by responding inelastically without losing significant vertical load-carrying capacity. C12.13 FOUNDATION DESIGN C12.13.3 Foundation Load-Deformation Characteristics. This section of the standard provides guidance on modeling load-deformation characteristics of the foundation-soil system (foundation stiffness) for linear analysis procedures. The further guidance contained herein addresses both linear and nonlinear analysis methods. Where linear analysis procedures are used with the methodology given below, the earthquake forces should not be reduced by R. Modeling of the load-deformation characteristics of foundations should be in accordance with ASCE/SEI 41. For nonlinear analysis of piles that may form plastic hinges, the lateral load-deformation characteristics of piles may be taken from Song, et al. (2005). For load combinations including seismic load effects, the vertical, lateral, and rocking load capacities of foundations as limited by the soil should be sufficient to resist loads with acceptable deformations, considering the short duration of loading, the dynamic properties of the soil, and the ultimate load capacities, Qus, of the foundations. Ultimate foundation load capacities should be determined by a qualified geotechnical engineer based on geotechnical site investigations that include field and laboratory testing to determine soil classification and soil strength parameters or on insitu testing of prototype foundations. For competent soils that do not undergo strength degradation under seismic loading, strength parameters for static loading conditions may be used to compute ultimate load capacities for seismic design. For sensitive cohesive soils or saturated cohesionless soils, the potential for earthquake-induced strength degradation should be considered. Ultimate foundation load capacities, Qus, under vertical, lateral, and rocking loading should be determined using accepted foundation design procedures and principles of plastic analysis. Calculated ultimate load capacities, Qus, should be bestestimated values using soil properties that are representative average values for individual foundations. Best-estimated values of Qus should be reduced by resistance factors (f) to reflect uncertainties in site conditions and in the reliability of analysis methods. The factored foundation load capacity, fQus, should be used both to check acceptance criteria and as the foundation capacity in nonlinear load-deformation models. If ultimate foundation load capacities are determined based on geotechnical site investigations including laboratory or in-situ tests, f factors equal to 0.8 for cohesive soils and 0.7 for cohesionless soils should be used for vertical, lateral, and rocking resistance for all foundation types. If ultimate foundation load capacities are determined based on full-scale field-testing of prototype foundations, f factors equal to 1.0 for cohesive soils and 0.9 for cohesionless soils are recommended. For both linear and nonlinear analysis procedures, a model incorporating a combined superstructure and foundation system is necessary to assess the effect of foundation deformations on the superstructure elements. For linear analysis methods, the linear load-deformation behavior of foundations should be represented by an equivalent linear (secant) stiffness using soil properties that are compatible with the soil strain levels associated with the design earthquake motion. The strain-compatible shear modulus, G, and the associated strain-compatible shear wave velocity, vs, needed for the evaluation of equivalent linear stiffness are specified in Chapter 19 of the standard or can be based on a sitespecific study. ASCE/SEI 41 is an acceptable alternative to that contained in the standard and may provide more realistic results. For nonlinear analysis procedures, the nonlinear load-deformation behavior of the foundation-soil system may be represented by a bilinear or multilinear curve having an initial equivalent linear stiffness and a limiting foundation capacity. The initial equivalent linear stiffness should be determined as described above for linear analysis methods. The limiting foundation capacity should be taken as the factored foundation load capacity, fQus. Parametric variations in analyses should include: (a) a reduction in stiffness of 50 percent combined with a limiting foundation capacity, f Qus, and (b) an increase in stiffness of 50 percent combined with a limiting foundation capacity equal to Qus multiplied by 1/f. For linear analysis procedures, factored foundation load capacities, f Qus, should not be exceeded for load combinations that include seismic load effects. For the nonlinear analysis procedures, if the factored foundation load capacity, f Qus, is reached during seismic loading, the potential significance of associated transient and permanent foundation displacements should be evaluated. Foundation displacements are acceptable if they do not impair the continuing function of Occupancy Category IV structures or the life safety of any structure. C12.13.4 Reduction of Foundation Overturning. Since the vertical distribution of forces prescribed for use with the equivalent lateral force procedure is intended to envelope story shears, overturning moments are exaggerated. (See Section C12.13.3.) Such moments will be lower where multiple modes respond, so a 25 percent reduction is permitted for design (strength and stability) of the foundation using this procedure. This reduction is not permitted for inverted pendulum or cantilevered column type structures, which typically have a single mode of response. Since the modal response spectrum analysis procedure more accurately reflects the actual distribution of shears and overturning moments, the permitted reduction is only 10 percent. C12.13.5 Requirements for Structures Assigned to Seismic Design Category C. C12.13.5.1 Pole-Type Structures. The high contact pressures that develop between pole and soil as a result of lateral loads make pole-type structures sensitive to earthquake motions. Bending in the poles and soil lateral capacity and deformation are key considerations in the design. For further discussion of pole-soil interaction, see Section C12.13.6.7. C12.13.5.2 Foundation Ties. One important aspect of adequate seismic performance is that the foundation acts as a unit, not permitting one column or wall to move appreciably with respect to another. To attain this performance, it is common to provide ties between footings and pile caps. This is especially important where the use of deep foundations is driven by the existence of soft surface soils. Multistory buildings often have major columns that run the full height of the building adjacent to smaller columns that support only one level; the calculated tie force is based on the heavier column load. The standard permits alternate methods of tying foundations together. Lateral soil pressure on pile caps is not a recommended method because motion is imparted from soil to structure and during displacement under dynamic conditions. C12.13.5.3 Pile Anchorage Requirements. The pile anchorage requirements are intended to prevent brittle failures of the connection to the pile cap under moderate ground motions. Moderate ground motions can result in pile tension forces or bending moments that could compromise shallow anchorage embedment. Loss of pile anchorage could result in increased structural displacements from rocking, overturning instability, and loss of shearing resistance at the ground surface. A concrete bond to a bare steel pile section usually is unreliable, but connection by means of deformed bars properly developed from the pile cap into concrete confined by a circular pile section is permitted. C12.13.6 Requirements for Structures Assigned to Seismic Design Categories D through F. C12.13.6.1 Pole-Type Structures. See Section C12.13.5.1. C12.13.6.2 Foundation Ties. See Section C12.13.5.2. For Seismic Design Categories D through F, the requirement is extended to spread footings on soft soils. C12.13.6.3 General Pile Design Requirements. Design of piles is based on the same R factor used in design of the superstructure; since inelastic behavior will result, piles should be designed with ductility similar to that of the superstructure. When strong ground motions occur, inertial structure pile-soil interaction may produce plastic hinging in piles near the bottom of the pile cap, and kinematic soil-pile interaction will result in bending moments and shearing forces throughout the length of the pile, being higher at interfaces between stiff and soft soil strata. These effects are particularly severe in soft soils and liquefiable soils so Section 14.2.3.2.1 requires special detailing in areas of concern. The shears and curvatures in piles caused by inertial and kinematic interaction may exceed the bending capacity of conventionally designed piles, resulting in severe damage. Analysis techniques to evaluate pile bending are discussed by Margason and Holloway (1977) and Mylonakis (2001), and these effects on concrete piles are further discussed by Shepard (1983). For homogeneous, elastic media and assuming the pile follows the soil, the free-field curvature (soil strains without a pile present) can be estimated by dividing the peak ground acceleration by the square of the shear wave velocity of the soil; considerable judgment is necessary in using this simple relationship for a layered, inelastic profile with pile-soil interaction effects. Norris (1994) discusses methods to assess pile-soil interaction. Where determining the extent of special detailing, the designer must consider variation in soil conditions and driven pile lengths, so that adequate ductility is provided at potential high curvature interfaces. Confinement of concrete piles to provide ductility and to maintain functionality of the confined core pile during and after the earthquake may be obtained by use of heavy spiral reinforcement or use of exterior steel liners. C12.13.6.4 Batter Piles. Partially embedded batter piles have a history of poor performance in strong ground shaking, as shown by Gerwick and Fotinos (1992). Failure of battered piles has been attributed to design that neglect loading on the piles from ground deformation or that assumes that lateral loads are resisted by axial response of piles without regard to moments induced in the pile at the pile cap (Lam and Bertero, 1990). Because batter piles are considered to have limited ductility, they must be designed using the load combinations with overstrength. Moment-resisting connections between pile and pile cap must resolve the eccentricities inherent in batter pile configurations. This concept is illustrated clearly by EQE Engineering (1991). C12.13.6.5 Pile Anchorage Requirements. Piles should be anchored to the pile cap to permit energy dissipating mechanisms, such as pile slip at the pile-soil interface, while maintaining a competent connection. This section of the standard sets forth a capacity design approach to achieve that objective. Anchorages occurring at pile cap corners and edges should be reinforced to preclude local failure of plain concrete sections due to pile shears, axial loads, and moments. C12.13.6.6 Splices of Pile Segments. A capacity design approach, similar to that for pile anchorage, is applied to pile splices. C12.13.6.7 Pile Soil Interaction. Short piles and long slender piles embedded in the earth behave differently when subject to lateral forces and displacements. The response of a long slender pile depends on its interaction with the soil considering the nonlinear response of the soil. Numerous design aid curves and computer programs are available for this type of analysis, which is necessary to obtain realistic pile moments, forces, and deflections and is common in practice (Ensoft, 2004). More sophisticated models, which also consider inelastic behavior of the pile itself, can be analyzed using general-purpose nonlinear analysis computer programs or closely approximated using the pile-soil limit state methodology and procedure given by Song, et al. (2005). Short piles (with length-to-diameter ratios no more than 6) can be treated as a rigid body simplifying the analysis. A method assuming a rigid body and linear soil response for lateral bearing is given in the current building codes. A more accurate and comprehensive approach using this method is given in a study by Czerniak (1957). C12.13.6.8 Pile Group Effects. The effects of groups of piles, where closely spaced, must be taken into account for vertical and horizontal response. As groups of closely spaced piles move laterally, failure zones for individual piles overlap, and horizontal strength and stiffness response of the pile-soil system is reduced. Reduction factors or “p-multipliers” are used to account for these groups of closely spaced piles. For a pile center-to-center spacing of three pile diameters, reduction factors of 0.6 for the leading pile row and 0.4 for the trailing pile rows are recommended by Rollins, et al. (1999). Computer programs are available to analyze group effects assuming a nonlinear soil and elastic piles (Ensoft, 2004a). C12.14 SIMPLIFIED ALTERNATIVE STRUCTURAL DESIGN CRITERIA FOR SIMPLE BEARING WALL OR BUILDING FRAME SYSTEMS C12.14.1 General. In recent years, engineers and building officials have become concerned that the seismic design requirements in codes and standards, while intended to make structures perform more reliably, have become so complex and difficult to understand and to implement that they may be counterproductive. Since the response of buildings to earthquake ground shaking is very complex (especially for irregular structural systems), realistically accounting for these effects can lead to complex requirements. There is a concern that the typical designers of small, simple structures, which may represent more than 90 percent of construction in the United States, have difficulty understanding and applying the general seismic requirements of the standard. The simplified procedure presented in this section of the standard applies to low-rise, stiff structures. The procedure, which was refined and tested over a five-year period, was developed to be used for a defined set of structures deemed to be sufficiently regular in configuration to allow a reduction of prescriptive requirements. For some design elements, such as foundations and anchorage of nonstructural systems, other sections of the standard must be followed, as referenced within Section 12.14. C12.14.1.1 Simplified Design Procedure. Reasons for the limitations of the simplified design procedure of Section 12.14 are as follows: 1. The procedure was developed to address adequate seismic performance for standard occupancies. Since it was not developed for higher levels of performance associated with Occupancy Category III and IV structures, no importance factor (I) is employed. 2. Site Class E and F soils require specialized procedures that are beyond the scope of the procedure. 3. The procedure was developed for stiff, low-rise buildings, where higher-mode effects are negligible. 4. Only stiff systems, where drift is not a controlling design criterion, may employ the procedure. Because of this limitation, drifts are not computed. The response modification coefficient, R, and the associated system limitations are consistent with those found in the general Chapter 12 requirements. 5. In order to achieve a balanced design and to achieve a reasonable level of redundancy, two lines of resistance are required in each of the two major axis directions. Because of this stipulation, no redundancy factor (.) is applied. 6. To reduce the potential for dominant torsional response, at least one line of resistance must be placed on each side of the center of mass. 7. Large overhangs for flexible diaphragm buildings can produce response that is inconsistent with the assumptions associated with the procedure. 8. A system that satisfies these layout and proportioning requirements avoids torsional irregularity, so calculation of accidental torsional moments is not required. 9. An essentially orthogonal orientation of lines of resistance effectively uncouples response along the two major axis directions, so orthogonal effects may be neglected. 10. Where the simplified design procedure is chosen, it must be used for the entire design, in both major axis directions. 11. Since in-plane and out-of-plane offsets generally create large diaphragm, collector, and discontinuous element demands that are not addressed by the procedure, these irregularities are prohibited. 12. Buildings that exhibit weak-story behavior violate the assumptions used to develop the procedure. C12.14.3 Seismic Load Effects and Combinations. The seismic load effect and combination equations for the simplified design procedure are consistent with those for the general procedure, with one notable exception: the overstrength factor (corresponding to O0 in the general procedure) is set at 2.5 for all systems as indicated in Section 12.14.3.2.1. Given the limited systems that can use the simplified design procedure, specifying unique overstrength factors was deemed unnecessary. C12.14.7 Design and Detailing Requirements. The design and detailing requirements outlined in this section are similar to those for the general procedure. The few differences include the following: 1. Forces used to connect smaller portions of a structure to the remainder of the structures are taken as 0.20 times the shortperiod design spectral response acceleration, SDS, rather than the general procedure value of 0.133 (Section 12.14.7.1). 2. Anchorage forces for concrete or masonry structural walls for structures with diaphragms that are not flexible are computed using the requirements for nonstructural walls. C12.14.8 Simplified Lateral Force Analysis Procedure C12.14.8.1 Seismic Base Shear. The seismic base shear in the simplified design procedure, as given by Equation 12.14-11, is a function of the short-period design spectral response acceleration, SDS. The value for F in the base shear equation addresses changes in dynamic response for two- and three-story buildings. As in the general procedure (Section 12.8.1.3), SDS may be computed for short, regular structures with SS taken no greater than 1.5. C12.14.8.2 Vertical Distribution. The seismic forces for multistory buildings are distributed vertically in proportion to the weight of the respective floor. Given the slightly amplified base shear for multi-story buildings, this assumption, along with the three-story height limit, produces results consistent with the more traditional triangular distribution without introducing that more sophisticated approach. C12.14.8.5 Drift Limits and Building Separation. For the simplified design procedure, which is restricted to stiff wall and braced frame structures, drift need not be calculated. Where drifts are required (such as for structural separations and cladding design) a conservative drift value of 1 percent is specified. REFERENCES American Society of Civil Engineers. 2006. Seismic Rehabilitation of Existing Buildings, ASCE/SEI 41. ASCE, Reston, Virginia. Applied Technology Council. 2009. Quantification of Building Seismic Performance Factors, FEMA P-695. Federal Emergency Management Agency, Washington, D.C. Applied Technology Council. 1978. Tentative Provisions for the Development of Seismic Regulations for Buildings, ATC 3- 06. ATC, Redwood City, California. Bernal, D. 1987. “Amplification Factors for Inelastic Dynamic P-delta Effects in Earthquake Analysis,” Earthquake Engineering and Structural Dynamics, 18: 635-681. Building Seismic Safety Council. 2004. NEHRP Recommended Provisions for Seismic Regulations for New Buildings and Other Structures, FEMA 450. Federal Emergency Management Agency, Washington, D.C. Charney, F. A. 1990. “Wind Drift Serviceability Limit State Design of Multistory Buildings,” Journal of Wind Engineering and Industrial Aerodynamics, 36:203-212. Charney, F. A., and J. Marshall. 2006. “A Comparison of the Krawinkler and Scissors Models for Including Beam-Column Joint Deformations in the Analysis of Moment-Resisting Frames,” AISC Engineering Journal, 43(1):31-48. Chopra, A. K. 2007. Structural Dynamics. Prentice Hall. Chopra, A. K., and R. K. Goel. 1991. “Evaluation of torsional provisions in seismic codes, Journal of Structural Engineering, 117(12):3762-3782 Chopra, A. K., and N. M. Newmark. 1980. “Analysis,” Chapter 2 of Design of Earthquake Resistant Structures, edited by E. Rosenblueth. John Wiley and Sons. Czerniak, E. 1957. "Resistance to Overturning of Single Short Piles" ASCE Journal of the Structural Division, 83(ST-2). Degenkolb, Henry J. 1987. “A Study of the P-delta Effect,” Earthquake Spectra, 3(1). De La Liera, J. C., and A. K. Chopra. 1994. “Evaluation of Code Accidental Torsion Records from Building Records,” Journal of Structural Engineering, 120(2):697-616. EQE Engineering. 1991. Structural Concepts and Details for Seismic Design, UCRL-CR-106554. Department of Energy, Washington, D.C. Ensoft, Inc. 2004a. Computer Program GROUP, Version 6.0, A Program for the Analysis of a Group of Piles Subjected to Axial and Lateral Loading, User’s Manual and Technical Manual. Ensoft, Austin, Texas. Ensoft, Inc. 2004b. Computer Program LPILE Plus Version 5.0, A Program for the Analysis of Piles and Drilled Shafts under Lateral Loads, User’s Manual and Technical Manual. Ensoft, Austin, Texas. Gerwick, Jr., B., and G. Fotinos. 1992. "Drilled Piers and Driven Piles for Foundations in Areas of High Seismicity," SEAONC Fall Seminar, October 29, San Francisco, California. Griffis, Larry. 1993. "Serviceability Limit States Under Wind Load," Engineering Journal, American Institute of Steel Construction, First Quarter. Gupta, A., and H. Krawinkler. 2000. "Dynamic P-delta Effects for Flexible Inelastic Steel Structures," Journal of Structural Engineering, 126(1):145-154. Lam, I., and V. Bertero. 1990. “Aseismic Design of Pile Foundations for Port Facilities,” in Proceedings of the POLA Seismic Workshop on Seismic Engineering, March 21-23, San Pedro, California, Port of Los Angeles. Liew, J. Y. 2001. “Inelastic Analysis of Steel Frames with Composite Beams,” Journal of Structural Engineering, 127(2):194-202. Lopez, O. A., and M. Cruz. 1996. "Number of Modes for the Seismic Design of Buildings," Earthquake Engineering and Structural Dynamics, 25(8):837-856. Margason, E., and M. Holloway. 1977. “Pile Bending During Earthquakes,” in Proceedings of the 6th World Conference on Earthquake Engineering, New Delhi. Mylonakis, G. 2001. “Seismic pile bending at soil-layer interfaces,” Soils and Foundations, 41 (4), pp. 47-58. Newmark, N. A., and E. Rosenbleuth. 1971. Fundamentals of Earthquake Engineering. Prentice Hall. Norris, G. M. 1994. “Seismic Bridge Pile Foundation Behavior,” in Proceedings, International Conference on Design and Construction of Deep Foundations, Federal Highway Administration, Vol. 1. Paulay, T. 1997. “Are Existing Seismic Torsion Provisions Achieving Design Aims?” Earthquake Spectra, 13(2):259-280. Paulay, T., and M. J. N. Priestly. 1992. Seismic Design of Reinforced Concrete and Masonry Structures. John Wiley and Sons. Rollins, Kyle M., K. T. Peterson, T. J. Weaver, and Andrew E. Sparks. 1999. “Static and dynamic lateral load behavior on a full-scale pile group in clay,” Brigham Young University, Provo, Utah, and the Utah Department of Transportation, Salt Lake City, June 23. Schaffhausen, R., and A. Wegmuller. 1977. “Multistory Rigid Frames with Composite Girders under Gravity and Lateral Forces,” AISC Engineering Journal, 2nd Quarter. Seismology Committee, Structural Engineers Association of California. 1996. Recommended Lateral Force Requirements and Commentary. SEAOC, Sacramento, California. Shepard, D. A. 1983. “Seismic Design of Concrete Piling,” PCI Journal (March/April). Song, S. T., Y. H. Chai, and T. H. Hale. 2005. “Analytical model for ductility assessment of fixed-head concrete piles,” ASCE Journal of Structural Engineering 131(7):1051-1059. Vamvatsikos, D. 2002. “Seismic Performance, Capacity and Reliability of Structures as Seen Through Incremental Dynamic Analysis,” Ph.D. Dissertation, Department of Civil and Environmental Engineering, Stanford University, Palo Alto, California. Wilson, E. L., A. Der Kiureghian, and E. P. Bayo. 1981. "A Replacement for the SRSS Method in Seismic Analysis," Earthquake Engineering and Structural Dynamics, Vol. 9. Wilson, E. L. 2000. Three-dimensional Static and Dynamic Analysis of Structures, Computers and Structures, Inc., Berkeley, California. Figure C13.1-1 Image of Hospital imaging equipment that fell from overhead mounts. COMMENTARY CHAPTER 13, SEISMIC DESIGN REQUIREMENTS FOR NONSTRUCTURAL COMPONENTS C13.1 GENERAL Chapter 13 defines minimum design criteria for architectural, mechanical, electrical, and other nonstructural systems and components recognizing structure use, occupant load, the need for operational continuity, and the interrelation of structural and architectural, mechanical, electrical, and other nonstructural components. Nonstructural components are designed for design earthquake ground motions as defined in Section 11.2 and determined in Section 11.4.4 of the standard. In contrast to structures, which are implicitly designed for a low probability of collapse when subjected to the maximum considered earthquake (MCE) ground motions, there are no implicit performance goals associated with the MCE for nonstructural components. Performance goals associated with the design earthquake are discussed in Section C13.1.3. Suspended or attached nonstructural components that could detach either in full or in part from the structure during an earthquake are referred to as falling hazards and may represent a serious threat to property and life safety. Critical attributes that influence the hazards posed by these components include their weight, their attachment to the structure, their failure or breakage characteristics (e.g., certain types of glass), and their location relative to occupied areas (e.g., over an entry or exit, a public walkway, an atrium, or a lower adjacent structure). Architectural components that pose potential falling hazards include parapets, cornices, canopies, marquees, glass, large ornamental elements (e.g., chandeliers), and building cladding. In addition, suspended mechanical and electrical components (e.g., mixing boxes, piping, and ductwork) may represent serious falling hazards. Figures C13.1-1 through C13.1-4 show damage to nonstructural components in past earthquakes. Figure C13.1-1 Hospital imaging equipment that fell from overhead mounts. Figure C13.1-2 Image of Collapsed light fixtures. Figure C13.1-3 Image of Collapsed duct and HVAC diffuser. Figure C13.1-2 Collapsed light fixtures. Figure C13.1-3 Collapsed duct and HVAC diffuser. Figure C13.1-4 Image of Damaged ceiling system. Figure C13.1-4 Damaged ceiling system. Components whose collapse during an earthquake could result in blockage of the means of egress deserve special consideration. The term “means of egress” is used commonly in building codes with respect to fire hazard. Consideration of egress may include intervening aisles, doors, doorways, gates, corridors, exterior exit balconies, ramps, stairways, pressurized enclosures, horizontal exits, exit passageways, exit courts, and yards. Items whose failure could jeopardize the means of egress include walls around stairs and corridors, veneers, cornices, canopies, heavy partition systems, ceilings, architectural soffits, light fixtures, and other ornaments above building exits or near fire escapes. Examples of components that generally do not pose a significant falling hazard include fabric awnings and canopies. Architectural, mechanical, and electrical components that, if separated from the structure, will fall in areas that are not accessible to the public (e.g., into a mechanical shaft or light well) also pose little risk to egress routes. For some architectural components such as exterior cladding elements, wind design forces may exceed the calculated seismic design forces. Nevertheless, seismic detailing requirements may still govern the overall structural design. Where this is a possibility, it must be investigated early in the structural design process. The seismic design of nonstructural components may involve consideration of nonseismic requirements that are affected by seismic bracing. For example, accommodation of thermal expansion in pressure piping systems often is a critical design consideration and seismic bracing for these systems must be arranged in a manner that accommodates thermal movements. Particularly in the case of mechanical and electrical systems, the design for seismic loads should not compromise the functionality, durability, or safety of the overall design; this requires collaboration between the various disciplines of the design and construction team. For various reasons (e.g., business continuity), it may be desirable to consider higher performance than that required by the building code. For example, to achieve continued operability of a piping system, it is necessary to prevent unintended operation of valves or other inline components in addition to preventing collapse and providing leak tightness. Higher performance also is required for components containing substantial quantities of hazardous contents (as defined in Section 11.2). These components must be designed to prevent uncontrolled release of those materials. The requirements of Chapter 13 are intended to apply to new construction and tenant improvements installed at any time during the life of the structure, provided they are listed in Table 13.5-1 or 13.6-1. Further, they are intended to reduce (not eliminate) the risk to occupants and to improve the likelihood that essential facilities remain functional. While property protection (in the sense of investment preservation) is a possible consequence of implementation of the standard, it is not currently a stated or implied goal; a higher level of protection may be advisable if such protection is desired or required. C13.1.1 Scope. The requirements for seismic design of nonstructural components apply to the nonstructural component as well as to its supports and attachments to the main structure. In some cases as defined in Section 13.2, it is necessary to consider explicitly the performance characteristics of the component. The requirements are intended to apply only to Figure C13.1-5 Image of Toppled storage cabinets. permanently attached components – not to furnishings, temporary items, or mobile units. Furnishings such as tables, chairs, and desks may shift during strong ground shaking but generally pose minimal hazards provided they do not obstruct emergency egress routes. Storage cabinets, tall bookshelves, and other items of significant mass do not fall into this category and should be anchored or braced in accordance with this chapter. Figure C13.1-5 Toppled storage cabinets. Temporary items are those that remain in place for short periods of time (months, not years). Components that, while movable or relocatable, are expected to remain in place for periods of a year or longer should be considered permanent for the purposes of this section. Modular office systems are considered permanent since they ordinarily remain in place for long periods. In addition, they often include storage units of significant capacity which may topple in an earthquake. They are subject to the provisions of Section 13.5.8 for partitions if they exceed 6 feet in height. Mobile units include components that are moved from one point in the structure to another during ordinary use. Examples include desktop computers, office equipment, and other components that are not permanently attached to the building utility systems (Figure C13.1-5). Components that are mounted on wheels to facilitate periodic maintenance or cleaning but that otherwise remain in the same location (e.g., server racks) are not considered moveable for the purposes of anchorage and bracing. Likewise, skid-mounted components (as shown in Figure C13.1-6) as well as the skids themselves are considered permanent equipment. In all cases, equipment must be anchored if it is permanently attached to utility services (electricity, gas, and water). For the purposes of this requirement, “permanently attached” should be understood to include all electrical connections except NEMA 5-15 and 5-20 straight-blade connectors (duplex receptacles). C13.1.2 Seismic Design Category. The requirements for nonstructural components are based in part on the Seismic Design Category to which they are assigned. As the Seismic Design Category is established considering factors not unique to specific nonstructural components, all nonstructural components occupying or attached to a structure are assigned to the same Seismic Design Category as the structure. C13.1.3 Component Importance Factor. Performance expectations for nonstructural components often are defined in terms of the functional requirements of the structure to which the components are attached. While specific performance goals for nonstructural components have yet to be defined in building codes, the component importance factor (Ip) implies performance levels for specific cases. For noncritical nonstructural components (those with an importance factor, Ip, of 1.0) the following behaviors are anticipated for shaking having different levels of intensity: 1. Minor earthquake ground motions – minimal damage; not likely to affect functionality 2. Moderate earthquake ground motions – some damage that may affect functionality 3. Design earthquake ground motions – major damage but significant falling hazards are avoided; likely loss of functionality. Figure C13.1-6 Image of Skid-mounted components. Figure C13.1-6 Skid-mounted components. Components with importance factors greater than 1.0 are expected to remain in place, sustain limited damage, and, when necessary, function following an earthquake (see Section C13.2.2). These components can be located in structures that are not assigned to Occupancy Category IV. For example, fire sprinkler piping systems have an importance factor, Ip, of 1.5 in all structures since these essential systems should function following an earthquake. The component importance factor is intended to represent the greater of the life-safety importance of the component and the hazard-exposure importance of the structure. It indirectly influences the survivability of the component via required design forces and displacement levels as well as component attachments and detailing. While this approach provides some degree of confidence in the seismic performance of a component, it may not be sufficient in all cases. For example, individual ceiling tiles may fall from a ceiling grid that has been designed for larger forces. This may not represent a serious falling hazard if the ceiling tiles are made of lightweight materials, but it may lead to blockage of critical egress paths or disruption of the facility function. When higher levels of confidence in performance are required, the component is classified as a designated seismic system (Section 11.2), and, in certain cases, seismic qualification of the component or system is necessary. Seismic qualification approaches are provided in Sections 13.2.5 and 13.2.6. In addition, seismic qualification approaches presently in use by the Department of Energy (DOE) can be applied. Occupancy Category IV structures are intended to be functional following a design earthquake; critical nonstructural components and equipment in such structures are designed with Ip equal to 1.5. This requirement applies to most components and equipment since damage to vulnerable unbraced systems or equipment may disrupt operations following an earthquake even if they are not directly classified as essential to life safety. The nonessential/nonhazardous components themselves are not assessed, and requirements focus solely on the supports and attachments. UFC 3-310-04 has additional guidance for improved performance. C13.1.4 Exemptions. Some nonstructural components either possess inherent strength and stability, are subject to low-level earthquake demands (accelerations and relative displacements), or both. Since these nonstructural components and systems are expected to achieve the performance goals described earlier in this commentary without explicitly satisfying additional requirements, they are exempt from the requirements of Chapter 13. Chapter 13 does not apply to Seismic Design Category A due to its very low level of seismic hazard. (See Section C11.7.) With the exception of parapets supported by bearing walls or shear walls, all components in Seismic Design Category B are exempt due to the minimal level of seismic risk. Parapets are not exempt because experience has shown these items can fail and pose a significant falling hazard even at low shaking levels. Mechanical and electrical components in Seismic Design Category C with an importance factor (Ip) equal to 1.0 are exempt because they are subject to low levels of seismic hazard, they do not contain hazardous substances, and their function is not required to maintain life safety following an earthquake. Small components with Ip equal to 1.0 in Seismic Design Categories D, E, and F also are exempt since they do not contain hazardous substances and are not large enough to pose a life-safety hazard if they fall, slide, or topple. Failures of unbraced distribution systems at or near the point of connection to nonstructural components have been observed in past earthquakes. For this reason, flexible connections such as expansion loops, braided hose, or expansion joints are required to allow for the larger relative displacements associated with unbraced components. Note that the stiffness of flexible connections may be sensitive to internal pressure and length of the connection. C13.1.5 Applicability of Nonstructural Component and Requirements. At times, a nonstructural component should be treated as a nonbuilding structure. When the physical characteristics associated with a given class of nonstructural components vary widely, judgment is needed to select the appropriate design procedure and coefficients. For example, cooling towers vary from small packaged units with an operating weight of 2,000 pounds or less to structures the size of buildings. Consequently, design coefficients for the design of “cooling towers” are found both in Table 13.6-1 and Table 15.4-2. Small cooling towers are best designed as nonstructural components using the provisions of Chapter 13 while large ones are clearly nonbuilding structures that are more appropriately designed using the provisions of Chapter 15. Similar issues arise for other classes of nonstructural component (e.g., boilers and bins). Guidance on determining whether an item should be treated as a nonbuilding structure or nonstructural component for the purpose of seismic design is provided in Bachman and Dowty (2008). There are practical limits on the size of a component that can be qualified via shake table testing. Components too large to be qualified by shake table testing need to be qualified by a combination of structural analysis and qualification testing or empirical evaluation through a subsystem approach. Subsystems of a large, complex component (e.g., a very large chiller) can be qualified individually and the overall structural frame of the component evaluated by structural analysis Premanufactured modular mechanical units are considered nonbuilding structures supporting nonstructural components. The entire system (all modules assembled) can be shake table qualified or qualified separately as subsystems. Modular mechanical units house various nonstructural components that are subject to all the design requirements of Chapter 13. The specified weight limit for nonstructural components (25 percent relative to the combined weight of the structure and component) relates to the condition at which dynamic interaction between the component and the supporting structural system is potentially significant. Section 15.3.2 contains requirements for addressing this interaction in design. C13.1.6 Reference Documents. Professional and trade organizations have developed nationally recognized codes and standards for the design and construction of specific mechanical and electrical components. These documents provide design guidance for normal and upset (abnormal) operating conditions and for various environmental conditions. Some of these documents include earthquake design requirements in the context of the overall mechanical or electrical design. It is the intent of the standard that seismic requirements in referenced documents be used. The developers of these documents are familiar with the expected performance and failure modes of the components; however, the documents may be based on design considerations not immediately obvious to a structural design professional. For example, in the design of industrial piping, stresses due to seismic inertia forces typically are not added to those due to thermal expansion. There is a potential for misunderstanding and misapplication of reference documents for the design of mechanical and electrical systems. A registered design professional familiar with both the standard and the reference documents used should be involved in the review and acceptance of the seismic design. Even when reference documents for nonstructural components lack specific earthquake design requirements, mechanical and electrical equipment constructed in accordance with industry-standard reference documents have performed well historically when properly anchored. Nevertheless, it is expected that manufacturers of mechanical and electrical equipment will consider seismic loads in the design of the equipment itself even when not explicitly required by this chapter. While some reference documents provide requirements for seismic capacity appropriate to the component being designed, the seismic demands used in design may not be less than those specified in the standard. Specific guidance for selected mechanical and electrical components and conditions is provided in Section 13.6. C13.1.7 Reference Documents Using Allowable Stress Design. Many nonstructural components are designed using specifically developed reference documents that are based on allowable stress loads and load combinations and permit increases in allowable stresses for seismic loading. Although Section 2.4.1 of the standard does not permit increases in allowable stresses, Section 13.1.7 explicitly defines the conditions for their use in the design of nonstructural components. C13.2 GENERAL DESIGN REQUIREMENTS C13.2.1 Applicable Requirements for Architectural, Mechanical, and Electrical Components, Supports, and Attachments. Compliance with the requirements of Chapter 13 may be accomplished by project-specific design or by a manufacturer’s certification of seismic qualification of a system or component. In each case, the evidence of compliance is submitted to the authority having jurisdiction. When compliance is by manufacturer's certification, the items must be installed in accordance with the manufacturer’s requirements. Components addressed by the standard include